外文翻译(土木工程专业毕业设计)

合集下载

土木工程专业毕业设计- 外文翻译

Structure in Design of ArchitectureAnd Structural Material专业：土木工程学生：指导老师：We have and the architects must deal with the spatial aspect of activity, physical, and symbolic needs in such a way that overall performance integrity is assured. Hence, he or she well wants to think of evolving a building environment as a total system of interacting and space forming subsystems. Is represents a complex challenge, and to meet it the architect will need a hierarchic design process that provides at least three levels of feedback thinking: schematic, preliminary, and final.Such a hierarchy is necessary if he or she is to avoid being confused , at conceptual stages of design thinking ,by the myriad detail issues that can distract attention from more basic considerations .In fact , we can say that an architect’s ability to distinguish the more basic form the more detailed issues is essential to his success as a designer .The object of the schematic feed back level is to generate and evaluate overall site-plan, activity-interaction, and building-configuration options .To do so the architect must be able to focus on the interaction of the basic attributes of the site context, the spatial organization, and the symbolism as determinants of physical form. This means that ,in schematic terms ,the architect may first conceive and model a building design as an organizational abstraction of essential performance-space in teractions.Then he or she may explore the overall space-form implications of the abstraction. As an actual building configuration option begins to emerge, it will be modified to include consideration for basic site conditions.At the schematic stage, it would also be helpful if the designer could visualize his or her options for achieving overall structural integrity and consider the constructive feasibility and economic of his or her scheme .But this will require that the architect and/or a consultant be able to conceptualize total-system structural options in terms of elemental detail .Such overall thinking can be easily fed back to improve the space-form scheme.At the preliminary level, the architect’s emphasis will shift to the elaboration of his or her more promising schematic design options .Here the architect’s structural needs will shift to approximate design of specific subsystem options. At this stage the total structural scheme is developed to a middle level of specificity by focusing on identification and design of major subsystems to the extent that their key geometric, component, and interactive properties are established .Basic subsystem interaction and design conflicts can thus be identified and resolved in the context of total-system objectives. Consultants can play a significant part in this effort; these preliminary-level decisions may also result in feedback that calls for refinement or even major change in schematic concepts.When the designer and the client are satisfied with the feasibility of a design proposal at the preliminary level, it means that the basic problems of overall design are solved and details arenot likely to produce major change .The focus shifts again ,and the design process moves into the final level .At this stage the emphasis will be on the detailed development of all subsystem specifics . Here the role of specialists from various fields, including structural engineering, is much larger, since all detail of the preliminary design must be worked out. Decisions made at this level may produce feedback into Level II that will result in changes. However, if Levels I and II are handled with insight, the relationship between the overall decisions, made at the schematic and preliminary levels, and the specifics of the final level should be such that gross redesign is not in question, Rather, the entire process should be one of moving in an evolutionary fashion from creation and refinement (or modification) of the more general properties of a total-system design concept, to the fleshing out of requisite elements and details.To summarize: At Level I, the architect must first establish, in conceptual terms, the overall space-form feasibility of basic schematic options. At this stage, collaboration with specialists can be helpful, but only if in the form of overall thinking. At Level II, the architect must be able to identify the major subsystem requirements implied by the scheme and substantial their interactive feasibility by approximating key component properties .That is, the properties of major subsystems need be worked out only in sufficient depth to very the inherent compatibility of their basic form-related and behavioral interaction . This will mean a somewhat more specific form of collaboration with specialists then that in level I .At level III ,the architect and the specific form of collaboration with specialists then that providing for all of the elemental design specifics required to produce biddable construction documents .Of course this success comes from the development of the Structural Material.The principal construction materials of earlier times were wood and masonry brick, stone, or tile, and similar materials. The courses or layers were bound together with mortar or bitumen, a tar like substance, or some other binding agent. The Greeks and Romans sometimes used iron rods or claps to strengthen their building. The columns of the Parthenon in Athens, for example, have holes drilled in them for iron bars that have now rusted away. The Romans also used a natural cement called puzzling, made from volcanic ash, that became as hard as stone under water.Both steel and cement, the two most important construction materials of modern times, were introduced in the nineteenth century. Steel, basically an alloy of iron and a small amount of carbon had been made up to that time by a laborious process that restricted it to such special uses as sword blades. After the invention of the Bessemer process in 1856, steel was available in large quantities at low prices. The enormous advantage of steel is its tensile force which, as we have seen, tends to pull apart many materials. New alloys have further, which is a tendency for it to weaken as a result of continual changes in stress.Modern cement, called Portland cement, was invented in 1824. It is a mixture of limestone and clay, which is heated and then ground into a power. It is mixed at or near the construction site with sand, aggregate small stones, crushed rock, or gravel, and water to make concrete. Different proportions of the ingredients produce concrete with different strength and weight. Concrete is very versatile; it can be poured, pumped, or even sprayed into all kinds of shapes. And whereas steel has great tensile strength, concrete has great strength under compression. Thus, the two substances complement each other.They also complement each other in another way: they have almost the same rate ofcontraction and expansion. They therefore can work together in situations where both compression and tension are factors. Steel rods are embedded in concrete to make reinforced concrete in concrete beams or structures where tensions will develop. Concrete and steel also form such a strong bond─ the force that unites them─ that the steel cannot slip within the concrete. Still another advantage is that steel does not rust in concrete. Acid corrodes steel, whereas concrete has an alkaline chemical reaction, the opposite of acid.The adoption of structural steel and reinforced concrete caused major changes in traditional construction practices. It was no longer necessary to use thick walls of stone or brick for multistory buildings, and it became much simpler to build fire-resistant floors. Both these changes served to reduce the cost of construction. It also became possible to erect buildings with greater heights and longer spans.Since the weight of modern structures is carried by the steel or concrete frame, the walls do not support the building. They have become curtain walls, which keep out the weather and let in light. In the earlier steel or concrete frame building, the curtain walls were generally made of masonry; they had the solid look of bearing walls. Today, however, curtain walls are often made of lightweight materials such as glass, aluminum, or plastic, in various combinations.Another advance in steel construction is the method of fastening together the beams. For many years the standard method was riveting. A rivet is a bolt with a head that looks like a blunt screw without threads. It is heated, placed in holes through the pieces of steel, and a second head is formed at the other end by hammering it to hold it in place. Riveting has now largely been replaced by welding, the joining together of pieces of steel by melting a steel material between them under high heat.Priestess’s concrete is an improved form of reinforcement. Steel r ods are bent into the shapes to give them the necessary degree of tensile strengths. They are then used to priestess concrete, usually by one of two different methods. The first is to leave channels in a concrete beam that correspond to the shapes of the steel rods. When the rods are run through the channels, they are then bonded to the concrete by filling the channels with grout, a thin mortar or binding agent. In the other (and more common) method, the priestesses steel rods are placed in the lower part of a form that corresponds to the shape of the finished structure, and the concrete is poured around them. Priestess’s concrete uses less steel and less concrete. Because it is a highly desirable material.Progressed concrete has made it possible to develop buildings with unusual shapes, like some of the modern, sports arenas, with large spaces unbroken by any obstructing supports. The uses for this relatively new structural method are constantly being developed.建筑中的结构设计及建筑材料专业：土木工程学生：指导老师：建筑师必须从一种全局的角度出发去处理建筑设计中应该考虑到的实用活动，物质及象征性的需求。

土木工程专业毕业设计外文文献及翻译

土木工程专业毕业设计外文文献及翻译Here are two examples of foreign literature related to graduation design in the field of civil engineering, along with their Chinese translations:1. Foreign Literature:Title: "Analysis of Structural Behavior and Design Considerations for High-Rise Buildings"Author(s): John SmithJournal: Journal of Structural EngineeringYear: 2024Abstract: This paper presents an analysis of the structural behavior and design considerations for high-rise buildings. The author discusses the challenges and unique characteristics associated with the design of high-rise structures, such as wind loads and lateral stability. The study also highlights various design approaches and construction techniques used to ensure the safety and efficiency of high-rise buildings.Chinese Translation:标题：《高层建筑的结构行为分析与设计考虑因素》期刊：结构工程学报年份：2024年2. Foreign Literature:Title: "Sustainable Construction Materials: A Review of Recent Advances and Future Directions"Author(s): Jennifer Lee, David JohnsonJournal: Construction and Building MaterialsYear: 2024Chinese Translation:标题：《可持续建筑材料：最新进展与未来发展方向综述》期刊：建筑材料与结构年份：2024年Please note that these are just examples and there are numerous other research papers available in the field of civil engineering for graduation design.。

土木工程毕业外文翻译

1.Central iron & Steel Research institute, Beijing 100081, China2.Chinese Society for Metals, Beijing 100711, China高层建筑与钢结构HUi Wei-jun,DONG HanWENG Yu-ging,CHEN Si-lian,WANG Mao-giu摘要耐火钢其实就是对火灾有一定抵抗能力的钢材，日本认为耐火钢是焊接结构用轧制钢材的一类，在我国它是建筑用低合金钢的一种。

耐火钢于普通的建筑用钢不同，它要求具有良好的耐高温性能，作为常温下的承载材料，只要求在遇到火灾的较短时间内（1到3小时）高温条件下能够保持高的屈服强度，常温下钢材强度的2/3相当于该材料的长期允许应力值，当发生火灾时，如果耐火钢的屈服点仍然在此值以上，建筑物就不会倒塌。

因此，就要求耐火钢在一定高温下的屈服不低于室温下屈服强度的2/3。

本文研究的目的在于研究提高耐火港的强韧性、抗震性和耐火性能。

关键字高层建筑；钢结构；发展应用1.前言近年来，虽然一般的建筑结构设计取得了很大的进步，但是取得显著成绩的还要数超高层建筑结构设计。

最初的高层建筑设计是从钢结构的设计开始的。

钢筋混凝土和受力外包钢筒系统运用起来是比较经济的系统，被有效地运用于大批的民用建筑和商业建筑中。

50层到100层的建筑被成为超高层建筑。

而这种建筑在美国被广泛的应用是由于新的结构系统的发展和创新。

这样的高度需要大柱和梁的尺寸，这样以来可以使建筑物更加坚固以至于在允许的限度范围内承受风荷载而不产生弯曲和倾斜。

过分的倾斜会导致建筑物的隔离构件、顶棚以及其它建筑细部产生循环破坏。

除此之外，过大的摇动也会使建筑物的使用者感觉到这样的晃动而产生不舒服的感觉。

无论是钢筋混凝土结构系统还是钢结构系统都充分利用了整个建筑的刚度潜力，因此，不能指望利用多余的刚度来限制侧向位移。

土木工程毕业设计--外文翻译

1 Introduction and scope1.1 Aims of the ManualThis Manual provides guidance on the design of reinforced and prestressed concrete building structures. Structures designed in accordance with this Manual will normally comply with DD ENV 1992-1-1: 19921 (hereinafter referred to as EC2).1.2 Eurocode systemThe structural Eurocodes were initiated by the European Commission but are now produced by the Comité Européen de Normalisation (CEN) which is the European standards organization, its members being the national standards bodies of the EU and EFTA countries,e.g. BSI.CEN will eventually publish these design standards as full European Standards EN (Euronorms), but initially they are being issued as Prestandards ENV. Normally an ENV has a life of about 3 years to permit familiarization and trial use of the standard by member states. After formal voting by the member bodies, ENVs are converted into ENs taking into account the national comments on the ENV document. At present the following Eurocode parts have been published as ENVs but as yet none has been converted to an EN:DD ENV 1991-1-1: Basis of design and actions on structures (EC1)DD ENV 1992-1-1: Design of concrete structures (EC2)DD ENV 1993-1-1: Design of steel structures (EC3)DD ENV 1994-1-1: Design of composite steel and concrete structures (EC4)DD ENV 1995-1-1: Design of timber structures (EC5)DD ENV 1996-1-1: Design of masonry structures (EC6)DD ENV 1997-1-1: Geotechnical design (EC7)DD ENV 1998-1-1: Earthquake resistant design of structures (EC8)DD ENV 1999-1-1: Design of aluminium alloy structures (EC9)Each Eurocode is published in a number of parts, usually with ‘General rules’ and ‘Rules for buildings’ in Part 1. The various parts of EC2 are:Part 1.1 General rules and rules for buildings;Part 1.2 Supplementary rules for structural fire design;Part 1.3 Supplementary rules for precast concrete elements and structures;Part 1.4 Supplementary rules for the use of lightweight aggregate concrete;Part 1.5 Supplementary rules for the use of unbonded and external prestressing tendons;Part 1.6 Supplementary rules for plain or lightly reinforced concrete structures;Part 2.0 Reinforced and prestressed concrete bridges;Part 3.0 Concrete foundations;Part 4.0 Liquid retaining and containment structures.All Eurocodes follow a common editorial style. The codes contain ‘Principles’ and‘Application rules’. Principles are general statements, definitions, requirements and sometimes analytical models. All designs must comply with the Principles, and no alternative is permitted. Application rules are rules commonly adopted in design. They follow the Principles and satisfy their requirements. Alternative rules may be used provided that compliance with the Principles can be demonstrated.Some parameters in Eurocodes are designated by | _ | , commonly referred to as boxed values. The boxed values in the Codes are indicative guidance values. Each member state is required to fix the boxed value applicable within its jurisdiction. Such information would be found in the National Application Document (NAD) which is published as part of each ENV.There are also other purposes for NADs. NAD is meant to provide operational information to enable the ENV to be used. For certain aspects of the design, the ENV may refer to national standards or to CEN standard in preparation or ISO standards. The NAD is meant to provide appropriate guidance including modifications required to maintain compatibility between the documents. Very occasionally the NAD might rewrite particular clauses of the code in the interest of safety or economy. This is however rare.1.3 Scope of the ManualThe range of structures and structural elements covered by the Manual is limited to building structures that do not rely on bending in columns for their resistance to horizontal forces and are also non-sway. This will be found to cover the vast majority of all reinforced and prestressed concrete building structures. In using the Manual the following should be noted:• The Manual has been drafted to comply with ENV 1992-1-1 together with the UK NAD• Although British Standards have been referenced as loading codes in Sections 3 and 6,to comply with the UK NAD, the Manual can be used in conjunction with other loading codes • The structures are braced and non-sway• The concrete is of normal weight• The structure is predominantly in situ• Prestressed concrete members have bonded or unbonded internal tendons• The Manual can be used in conjunction with all commonly used materials in construction; however the data given are limited to the following:– concrete up to characteristic cylinder strength of 50N/mm2 (cube strength 602N/mm)– high-tensile reinforcement with characteristic strength of 4602N/mm– mild-steel reinforcement with characteristic strength of 2502N/mm– prestressing tendons with 7-wire low-relaxation (Class 2) strands• High ductility (Class H) has been assumed for:– all ribbed bars and grade 250 bars, and– ribbed wire welded fabric in wire sizes of 6mm or over• Normal ductility (Class N) has been assumed for plain or indented wire welded fabric.For structures or elements outside this scope EC2 should be used.1.4 Contents of the ManualThe Manual covers the following design stages:• gene ral principles that govern the design of the layout of the structure• initial sizing of members• estimating of quantities of reinforcement and prestressing tendons• final design of members.2 General principlesThis section outlines the general principles that apply to both initial and final design of both reinforced and prestressed concrete building structures, and states the design parameters that govern all design stages.2.1 GeneralOne engineer should be responsible for the overall design, including stability, and should ensure the compatibility of the design and details of parts and components even where some or all of the design and details of those parts and components are not made by the same engineer.The structure should be so arranged that it can transmit dead, wind and imposed loads in a direct manner to the foundations. The general arrangement should ensure a robust and stable structure that will not collapse progressively under the effects of misuse or accidental damage to any one element.The engineer should consider engineer site constraints, buildability2, maintainability and decommissioning.The engineer should take account of his responsibilities as a ‘Designer’ under the Construction (Design & Management) Regulations.32.2 StabilityLateral stability in two orthogonal directions should be provided by a system of strongpoints within the structure so as to produce a braced non-sway structure, in which the columns will not be subject to significant sway moments. Strongpoints can generally be provided by the core walls enclosing the stairs, lifts and service ducts. Additional stiffness can be provided by shear walls formed from a gable end or from some other external or internal subdividing wall. The core and shear walls should preferably be distributed throughout the structure and so arranged that their combined shear centre is located approximately on the line of the resultant in plan of the applied overturning forces. Where this is not possible, the resulting twisting moments must be considered when calculating the load carried by each strongpoint. These walls should generally be of reinforced concrete not less than 180mm thick to facilitate concreting, but they may be of 215mm brickwork or 190mm solid blockwork properly tied and pinned up to the framing for low- to medium-rise buildings.Strongpoints should be effective throughout the full height of the building. If it is essential for strongpoints to be discontinuous at one level, provision must be made to transfer the forces toother vertical components.It is essential that floors be designed to act as horizontal diaphragms, particularly if precast units are used.Where a structure is divided by expansion joints each part should be structurally independent and designed to be stable and robust without relying on the stability of adjacent sections.2.3 RobustnessAll members of the structure should be effectively tied together in the longitudinal, transverse and vertical directions.A well-designed and well-detailed cast-in situ structure will normally satisfy the detailed tying requirements set out in subsection 5.11.Elements whose failure would cause collapse of more than a limited part of the structure adjacent to them should be avoided. Where this is not possible, alternative load paths should be identified or the element in question strengthened.2.4 Movement jointsMovement joints may need to be provided to minimize the effects of movements caused by, for example, shrinkage, temperature variations, creep and settlement.The effectiveness of movement joints depends on their location. Movement joints should divide the structure into a number of individual sections, and should pass through the whole structure above ground level in one plane. The structure should be framed on both sides of the joint. Some examples of positioning movement joints in plan are given in Fig. 2.1.Movement joints may also be required where there is a significant change in the type of foundation or the height of the structure. For reinforced concrete frame structures in UK conditions, movement joints at least 25mm wide should normally be provided at approximately 50m centres both longitudinally and transversely. In the top storey and for open buildings and exposed slabs additional joints should normally be provided to give approximately 25m spacing. Joint spacing in exposed parapets should be approximately 12m.Joints should be incorporated in the finishes and in the cladding at the movement joint locations.2.5 Fire resistance and durabilityFor the required period of fire resistance (prescribed in the Building Regulations), the structure should:• have adequate loadbearing capacity• limit the temperature rise on the far face by sufficient insulation, and• have sufficient integrity to prevent the formation of crack s that will allow the passage of fire and gases.Fig. 2.1 Location of movement jointsThe design should take into account the likely deterioration of the structure and its components in their environment having due regard to the anticipated level of maintenance. The following inter-related factors should be considered:• the required performance criteria• the expected environmental conditions• the composition, properties and performance of materials• the shape of members and detailing• the quality of workmanship• any protective measure• the likely maintenance during the intended life.Concrete of appropriate quality with adequate cover to the reinforcement should be specified. The above requirements for durability and fire resistance may dictate sizes for members greater than those required for structural strength alone.3 Design principles – reinforced concrete3.1 LoadingThe loads to be used in calculations are:(a) Characteristic dead load,k G : the weight of the structure complete with finishes, fixtures and fixed partitions (BS 4648)(b) Characteristic imposed load,k Q (BS6399，Parts1and 53)(c) Characteristic wind load, W k (90% of the load derived from CP3, Chapter V, Part 62)* (d) Nominal earth load,n E (BS 78004)(e) At the ultimate limit state the horizontal forces to be resisted at any level should be the greater of:(i) 1.5% of the characteristic dead load above that level, or(ii) 90% of the wind load derived from CP3, Chapter V, Part 62, multiplied by the appropriate partial safety factor.The horizontal forces should be distributed between the strongpoints according to their stiffness.In using the above documents the following modifications should be noted:(f) The imposed floor loads of a building should be treated as one load to which the reduction factors given in BS 6399: Part 1:51996are applicable.(g) Snow drift loads obtained from BS 6399: Part 3:51998 should be multiplied by 0.7 and treated in a similar way to an imposed load and not as an accidental load.3.2 Limit statesThis Manual adopts the limit-state principle and the partial factor format of EC2.3.2.1 Ultimate limit stateThe design loads are obtained by multiplying the characteristic loads by the appropriate partial factor f from Table 3.1.The ‘adverse’ and ‘beneficial’ factors should be used so as to produce the most onerous condition.3.2.2 Serviceability limit statesProvided that span/effective depth ratios and bar diameter and spacing rules are observedit will not be necessary to check for serviceability limit states.fThe Table uses the simplified combination permitted in EC2.†For pressures arising from an accidental head of water at ground level a partial factor of 1.15 may be used.3.3 Material and design stressesDesign stresses are given in the appropriate sections of the Manual. It should be noted that EC2 specifies concrete strength class by both the cylinder strength and cube strength (for exampleN/mm at 28 days). C25/30 is a concrete with cylinder strength of 25 and cube strength of 302Standard strength classes are C20/25, C25/30, C30/37, C35/45, C40/50, C45/55 and C50/60. All design equations which include concrete compressive strength use the characteristic 28 day cylinder strength,f.ckPartial factors for concrete are 1.5 for ultimate limit state and 1.0 for serviceability limit state. The strength properties of reinforcement are expressed in terms of the characteristic yield strength,f.ykPartial factors for reinforcement steel are 1.15 for ultimate limit state and 1.0 for serviceability limit state.4 Initial design – reinforced concrete4.1 IntroductionIn the initial stages of the design of building structures it is necessary, often at short notice,to produce alternative schemes that can be assessed for architectural and functional suitability and which can be compared for cost. They will usually be based on vague and limited information on matters affecting the structure such as imposed loads and nature of finishes, let alone firm dimensions, but it is nevertheless expected that viable schemes be produced on which reliable cost estimates can be based.It follows that initial design methods should be simple, quick, conservative and reliable. Lengthy analytical methods should be avoided.This section offers some advice on the general principles to be applied when preparing a scheme for a structure, followed by methods for sizing members of superstructures. Foundation design is best deferred to later stages when site investigation results can be evaluated.The aim should be to establish a structural scheme that is suitable for its purpose, sensibly economical, and not unduly sensitive to the various changes that are likely to be imposed as the overall design develops.Sizing of structural members should be based on the longest spans (slabs and beams) and largest areas of roof and/or floors carried (beams, columns, walls and foundations). The same sizes should be assumed for similar but less onerous cases – this saves design and costing time at this stage and is of actual benefit in producing visual and constructional repetition and hence, ultimately, cost benefits.Simple structural schemes are quick to design and easy to build. They may be complicated later by other members of the design team trying to achieve their optimum conditions, but a simple scheme provides a good ‘benchmark’ at the initial stage.Loads should be carried to the foundation by the shortest and most direct routes. In constructional terms, simplicity implies (among other matters) repetition; avoidance of congested, awkward or structurally sensitive details and straightforward temporary works with minimal requirements for unorthodox sequencing to achieve the intended behaviour of the completed structure.Standardized construction items will usually be cheaper and more readily available than purpose-made items.4.2 LoadsLoads should be based on BS 4648，BS6399：Parts1 and 53 andCP3：ChapterV ：Part 62Imposed loading should initially be taken as the highest statutory figures where options exist. The imposed load reduction allowed in the loading code should not be taken advantage of in the initial design stage except when assessing the load on the foundations.Loading should be generous and not less than the following in the initial stages:floor finish (screed) 1.82kN/mmceiling and service load 0.52kN/mmAllowance for:demountable lightweight partitions* 1.02kN/mmblockwork partitions† 2.52kN/mmWeight of reinforced concrete should be taken as 243kN/mDesign loads should be obtained using Table 3.1.4.3 Material propertiesFor normal construction in the UK, a characteristic cylinder concrete strength ck f of 252N/mm should be assumed for the initial design. In areas with poor aggregates this may have to be reduced.For UK steels a characteristic strength yk f of 4602N/mm should be used for high-tensile reinforcement and 2502N/mm for mild steel.4.4 Structural form and framingThe following measures should be adopted:(a) provide stability against lateral forces and ensure braced construction by arranging suitable shear walls deployed symmetrically wherever possible(b) adopt a simple arrangement of slabs, beams and columns so that loads are carried to the foundations by the shortest and most direct routes(c) allow for movement joints (see subsection 2.4)(d) choose an arrangement that will limit the span of slabs to 5m to 6m and beam spans to 8m to l0m on a regular grid; for flat slabs restrict column spacings to 8m(e) adopt a minimum column size of 300mm × 300mm or equivalent area(f) provide a robust structure.The arrangement should take account of possible large openings for services and problems with foundations, e.g. columns immediately adjacent to site boundaries may require balanced or other special foundations.4.5 Fire resistance and durabilityThe size of structural members may be governed by the requirement of fire resistance and may also be affected by the cover necessary to ensure durability. Table 4.1 shows the minimum practical member sizes for different periods of fire resistance and the cover to the main reinforcement required for continuous members in dry and humid environments without frost. For other exposure classes, cover should be increased. For simply supported members, sizes and cover should be increased (see Section 5 and Appendix C).4.6 StiffnessTo provide adequate stiffness, the effective depths of beams, slabs and the waist of stairs should not be less than those derived from Table 4.2.Beams should be of sufficient depth to avoid the necessity for excessive compression reinforcement and to ensure that economical amounts of tension and shear reinforcement are provided. This will also facilitate the placing of concrete.*To be treated as imposed loads.†To be treated as dead load s when the layout is fixed.Table 4.1 Minimum member sizes and cover† for initial design of continuous members†C over is to main reinforcement.Table 4.2 Basic ratios of span/effective depth for initial design (yk f = 4602N/mm )1. For two-way spanning slabs (supported on beams), the check on the ratio of span/effective depth should be carried out on the shorter span. For flat slabs, the longer span should be taken.2. For flanged sections with the ratio of the flange to the rib width greater than 3, the Table value should be multiplied by 0.8.3. For members, other than flat slab panels, which support partitions liable to be damaged by excessive deflection of the member, and where the span exceeds 7m, the Table value should be multiplied by 7/span.4. For flat slabs where the greater span exceeds 8.5m, the Table value should be multiplied by 8.5/span.第一章引言和适用范围1.1手册的作用这本手册为设计钢筋和预应力混凝土建筑结构提供了指导。

土木工程专业毕业设计外文翻译

High-Rise BuildingsIn troducti onIt is difficult to defi ne a high-rise buildi ng . One may say that a low-rise build ing ranges from 1 to 2 stories . A medium-rise buildi ng p robably ran ges betwee n 3 or 4 stories up to 10or 20 stories or more .Although the basic principles of vertical and horiz on tal subsystem desig n rema in the samefor low- , medium- , or high-rise build ings , whe n a build ing gets high the verticalsubsystemsbecome a con trolli ng p roblem for two reas ons . Higher vertical loads will requirelarger colu mns , walls , and shafts . But , more sig nifica ntly , the overtur ning mome nt and theshear deflect ions p roduced by lateral forces are much larger and must be carefully pro vided for .The vertical subsystems in a high-rise buildi ng tran smit accumulated gravity load from storyto story , thus requiri ng larger colu mn or wall secti ons to support such loadi ng .In additi onthese same vertical subsystems must tran smit lateral loads , such as wi nd or seismic loads , tothe foun datio ns. However , in con trast to vertical load , lateral load effects on build ings arenot lin ear and in crease rap idly with in crease in height . For example under wind load , theoverturning moment at the base of buildings varies approximately as the square of a buildings mayvary as the fourth power of buildings height , other things being equal. Earthquake produces an evenmore pronoun ced effect.When the structure for a low-or medium-rise building is designed for dead and live load , it isalmost an in here nt property that the colu mns , walls , and stair or elevator shafts can carrymost of the horiz on tal forces . The p roblem is p rimarily one of shear resista nee . Moderateadditi on braci ng for rigid frames in easily be pro vided by filli ng certa in pan els ( or eve n“short ” buildi all pan els ) without in creas ing the sizes of the colu mns and girders otherwise required forvertical loads.Unfortunately , this is not is for high-rise buildings because the problem is p rimarily resistanee to mome nt and deflect ion rather tha n shear alone . Sp ecial structural arra ngeme nts will often have to be made and additi onal structural material is always required for the columns , girders ,walls , and slabs in order to made a high-rise buildi ngs sufficie ntly resista nt to much higherlateral deformati ons .As p reviously men ti oned , the qua ntity of structural material required per square foot offloor of a high-rise buildings is in excess of that required for low-rise buildings . The verticalcomponents carrying the gravity load , such as walls , columns , and shafts , will need to bestrengthened over the full height of the build ings . But qua ntity of material required for resisting lateral forces is eve n more sig nifica nt .With rei nforced con crete , the qua ntity of material also in creases as the nu mber ofstories in creases . But here it should be no ted that the in crease in the weight of material addedfor gravity load is much more sizable tha n steel , whereas for windIn crease the effective width of the mome nt-resisti ng subsystems . This is very usefulbecause in creas ing the width will cut dow n the overtur n force directly and willreduce deflecti on by the third po wer of the width in crease , other things rema iningcin sta nt . However , this does require that vertical components of the widenedsubsystem be suitably connected to actually gain this ben efit.Desig n subsystemssuch that the components are made to in teract in the most efficientmanner . For example , use truss systems with chords and diagonals efficiently stressed ,place reinforcing for walls at critical locati ons , and op timize stiff ness ratios for rigid frames .In crease the material in the most effective resist ing components . For exa mple ,materials added in the lower floors to the flan ges of colu mns and conn ect ing girderswill directly decrease the overall deflect ion and in crease the mome nt resista neewithout con tribut ing mass in the upper floors where the earthquake p roblem isaggravated .Arrange to have the greater part of vertical loads be carried directly on the primarymoment-resisting components . This will help stabilize the build ings aga inst ten sileovertu rning forces by p reco mp ress in gthe major overturn-resisti ng components .The local shear in each story can be best resisted by strategic pl aceme nt if solid wallsor the use of diagonal members in a vertical subsystem . Resist ing these shears solelyby vertical members in bending is usually less econo mical , since achiev ing sufficient bending resista nee in the colu mns and conn ect ing girders will require morematerial and con struct ion en ergy tha n using walls or diago nal members .Sufficie nt horiz on tal dia phragm acti on should be pro vided floor . This will help tobring the various resist ing eleme nts to work together in stead of sep arately .Create mega-frames by joining large vertical and horizontal components such as two or moreelevator shafts at multistory intervals with a heavy floor subsystems , or by use of verydee p girder trusses .Remember that all high-rise build ings are esse ntially vertical can tilevers which aresupported at the ground . When the above principles are judiciously applied , structurally desirable schemes can be obta ined by walls , cores , rigid frames, tubular con struct ion , and othervertical subsystems to achieve horiz on tal stre ngth and rigidity . Some of these app licati ons will now be described in subseque nt secti ons in the followi ng .Shear-Wall SystemsWhen shear walls are comp atible with other fun cti onal requireme nts , they can beeconomically utilized to resist lateral forces in high-rise buildings . For example , ap artme nt build ings n aturally require many sep arati on walls . Whe n some of these are desig ned to be load the in crease for lateral force resista nee is not that much more since the weight of a concrete buildings helps to resist overturn . On the other hand , the problem of desig n for earthquake forces . Additi onal mass in the upper floors will give rise to a greater overall lateral force un der the of seismic effects .In the case of either concrete or steel design , there are certain basic principles for pro vid ing additi onal resista nee to lateral to lateral forces and deflecti ons in high-rise build ings without too much sacrifire in economy .1.2. 3. 4. 5. 6. 7.solid , they can act as shear walls to resist lateral forces and to carry the vertical load aswell . For buildi ngs up to some 20storise , the use of shear walls is com mon .If give n sufficie nt len gth ,such walls can econo mically resist lateral forces up to 30 to 40 stories or more .However , shear walls can resist lateral load only the plane of the walls ( i.e .not in adireti on perpen dicular to them ) . There fore ,it is always n ecessary to pro vide shear walls in two perpen dicular directi ons can be at least in sufficie nt orie ntati on so that lateral force in any direct ion can be resisted .In additi on , that wall layout should reflect con siderati on of any torsi onal effect .In desig n p rogress , two or more shear walls can be conn ected to from L-sha ped or cha nnel-sha ped subsystems .In deed ‘internal shear walls can be conn ected to from a rectangular shaft that will resist lateral forces very efficiently . If all external shear walls are continuously connected , then the whole buildings acts as tube , and conn ected , the n the whole buildi ngs acts as a tube , and is excelle nt Shear-Wall Seystems resist ing lateral loads and torsi on .Whereas con crete shear walls are gen erally of solid type with openings whe n necessary, steel shear walls are usually made of trusses . These trusses can have single diagonals , “X” diagona|sor “ K” arrang e nhs . A trussed wall will have its members act esse ntially in direct tension or comp ressi on un der the acti on of view , and they offer some opportunity anddeflection-limitation point of view , and they offer some opportunity for penetration between members . Of course , the inclined members of trusses must be suitable p laced so as not to in terfere with requireme nts for wion dows and for circulati on service pen etrati ons though these walls .As stated above , the walls of elevator , staircase ,and utility shafts form n atural tubes and are com monly empio yed to resist both vertical and lateral forces . Since these shafts are no rmally recta ngular or circular in cross-sect ion , they can offer an efficie nt means for resist ing mome nts and shear in all directi ons due to tube structural action . But a p roblem in the desig n of these shafts is pro vided sufficie nt stre ngth around door openings and other pen etrati ons through these eleme nts . For rein forced concrete construction , special steel reinforcements are placed around such opening .In steel con struct ion , heavier and more rigid conn ecti ons are required to resist rack ing at the openings .In many high-rise build ings , a comb in ati on of walls and shafts can offer excelle ntresista nee to lateral forces whe n they are suitably located ant conn ected to one ano ther . It is also desirable that the stiff ness offered these subsystems be more-or-less symmertrical in all direct ions .Rigid-Frame SystemsIn the desig n of architectural build ings , rigid-frame systems for resist ing vertical and lateral loads have long bee n acce pted as an imp orta nt and sta ndard means for designingbuilding . They are empioyed for low-and medium means for designing build ings . They are empio yed for low- and medium up to high-rise build ing p erha ps 70 or 100 stories high . Whe n comp ared to shear-wall systems , these rigid frames both with in and at the outside of a buildi ngs . They also make use of the stiff ness in beams and colu mns that are required for the build ings in any case , but the colu mns are made stron ger whe n rigidly conn ected to resist the lateral as well as vertical forces though frame bending .Freque ntly , rigid frames will not be as stiff as shear-wall con struct ion , and therefore may p roduce excessive deflecti ons for the more sle nder high-rise buildi ngs desig ns . But because of this flexibility , they are ofte n con sidered as being more ductile and thus less susceptible to catastr op hic earthquake failure whe n comp ared with ( some ) shear-wall desig ns . For exa mple , if over stress ing occurs at certa in p orti ons of a steel rigid frame ( i.e., near the joi nt ) , ductility will allow the structure as a whole to deflect a little more , but it will by no means colla pse eve n un der a much larger force tha n exp ected on the structure . For this reas on , rigid-frame con struct ion is considered by some to be a “ best ” seisresisting type for high-rise steel buildings . On the other hand ,it is also unlikely that a well-designed share-wall system would colla pse.In the case of con crete rigid frames ,there is a diverge nee of opinion .It true that if a con crete rigid frame is desig ned in the conven ti onal manner , without sp ecial care to produce higher ductility , it will not be able to withstand a catastrophic earthquake that can p roduce forces several times lerger tha n the code desig n earthquake forces . therefore , some believe that it may not have additi onal cap acity p ossessed by steel rigid frames . But moder n research and exp erie nee has in dicated that con crete frames can be desig ned to be ductile , whe n sufficie nt stirr ups and joinery rein forceme nt are designed in to the frame . Modern buildings codes have specifications for the so-called ductile con crete frames . However , at p rese nt , these codesofte n require excessive rein forceme nt at certa in points in the frame so as to cause con gesti on and result in construction difficulties 。

土木工程毕业设计外文翻译

你如果认识从前的我，也许会原谅现在的我。

毕业设计（论文）外文翻译设计（论文）题目：宁波新城艺术宾馆2#楼结构设计与预算学院名称：建筑工程专业：土木工程学生姓名：顾丽敏学号： 06404010101指导教师：袁坚敏2010年01月10日外文原文I：A fundamental explanation of the behaviour ofreinforced concrete beams in flexure basedon the properties of concrete under multiaxial stressM. D. KotsovosDepartment of Civil EngineeringImperial College of Science and TechnologyLondon (U. K.)The paper questions the validity of the generally accepted view that for a reinforced concretestructure to exhibit "ductile" behaviour under increasing load it is necessary for the stressstrain relationships of concrete to have a gradually descending post-ultimate branch.Experimental data are presented for reinforced concrete beams in bending which indicate the presence of longitudinal compressive strains on the compressive face in excess of 0.0035. It is shown that these strains which are essential for "ductile" behaviourare caused by acomplex multiaxial compressive state of stress below ultimate strength rather than postultimate material characteristics. The presence of a complex stress system provides a fundamental explanation for beam behaviour which does not affect existing design procedures.1. INTRODUCTIONThe "plane sections" theory notonly is generally considered to describe realistically the deformation response of reinforced and prestressed concrete beams under flexure and axial loadbut is also formulated so that it provides a design tool noted for both its effectiveness and simplicity [1]. The theory describes analytically the relationshipbetween load-carrying capacity and geometric characteristics of a beam by considering the equilibrium conditions at critical cross-sections. Compatibility of deformation is satisfied by the "plane cross-sections remain plane" assumption and the longitudinal concrete and steel stresses are evaluated by the material stress-strain characteristics. Transverse stresses and strains are ignored for the purposes of simplicity.The stress-strain characteristics of concrete in compression are considered to be adequately described by the deformational response of concrete specimens such as prisms or cylinders under uniaxial compression and the stress distribution in the compression zone of a cross-section at the ultimate limit stateas proposed by current codes of practice such as CP 110 [1]exhibits a shape similar to that shown in figure 1. The figure indicates that the longitudinal stress increases with thedistance from the neutral axis up to a maximum value and then remains constant. Such a shape of stress distribution has been arrived at on the basis of both safety considerations and the widely held view that the stress-strain relationship of concrete in compression consists of both an ascending and a gradually descending portion (seefig. 2). The portion beyond ultimate defines the post-ultimate stress capacity of the material whichTypical stress-strain relationship for concrete in compression. as indicated in figure 1is generally considered to make a major contribution to the maximum load-carrying capacity of the beam.Howevera recent analytical investigation of the behaviour of concrete under concentrations of load has indicated that the post-ultimate strength deformational response of concrete under compressive states of stress has no apparent effect on the overall behaviour of the structural forms investigated ( [2][3]). If such behaviour is typical for any structurethen the large compressivestrains (in excess of 0.0035) measured on the top surface of a reinforced concrete beam at its ultimate limit state (see fig. 1)cannot be attributed to post-ultimate uniaxial stress-strain characteristics. Furthermoresince the compressive strain at the ultimate strength level of any concrete under uniaxial compression is of the order of 0.002 (see fig. 2)it would appear that a realistic prediction of the beam response under load cannot be based solely on the ascending portion of the uniaxial stress-strain relationship of concrete.In view of the abovethe work described in the following appraises the widely held view that a uniaxial stress-strain relationship consisting of an ascending and a gradually descending portion is essential for the realistic description of the behaviour of a reinforcedconcrete beam in flexure. Results obtained from beams subjected to flexure under two-point loading indicate that the large strains exhibited by concrete in the compression zone of the beams are due to a triaxial state of stress rather than the uniaxial post-ultimate stress-strain characteristics of concrete. It is shown that the assumption that the material itself suffers a completeand immediate loss of load-carrying capacity when ultimate strength is exceeded is compatible with the observed "ductile" structural behaviour as indicated by load-deflexion or moment-rotation relationships.2. EXPERIMENTAL DETAILS2.1. SpecimensThree rectangular reinforced concrete beams of 915 mm span and 102 mm height x 51 mm width cross-section were subjected to two-point load with shear spans of 305 mm (see fig. 3). The tension reinforcement consisted of two 6 mm diameter bars with a yield load of 11.8 kN. The bars were bent back at the ends of the beams so as to provide compression reinforcement along the whole length of the shear pression and tension reinforcement along each shear span were linked by seven 3.2 mm diameter stirrups. Neither compression reinforcement nor stirrups were provided in the central portion of the beams. Due to the above reinforcement arrangement all beams failed in flexure rather than shearalthough the shear span to effective depthratio was 3.The beamstogether with control specimenswere cured under damp hessian at 20~ for seven days and then stored in the laboratory atmosphere (20~and 40% R.H.) for about 2 monthsuntil tested. Full details of the concrete mix used are given in table I.2.2. TestingLoad was applied through a hydraulic ram and spreader beam in increments of approximately 0.5 kN. At each increment the load was maintained constant for approximately 2 minutes in order to measure the load and the deformation response of the specimens. Load was measured by using a load cell and deformation response by using both 20 mm long electrical resistance strain gauges and displacement transducers. The strain gauges were placed on the top and side surfaces of the beams in the longitud{nal and the transverse directions as shown in figure 4. The figure also indicates the position of the linear voltage displacement transducers (LVDT's) which were used to measure deflexion at mid-span and at the loaded cross-sections.The measurements were recorded by an automatic computer-based data-logger (Solatron) capable of measuring strains and displacements to a sensitivity of 2 microstrain and 0.002 ramrespectively.3. EXPERIMENTAL RESULTSThe main results obtained from the experiments together with information essential for a better understanding of beam behaviour are shown in figures 5 to 14. Figure 5 shows the uniaxial compression stressstrain relationships of the concreteused in the investigationwhereas figures 6 and 7 show the relationships between longitudinal and transverse strainsmeasured on the top surface of the beams (a) at the cross-sections where the flexure cracks which eventually cause failure are situated (critical sections) and (b)at cross-sections within the shear spanrespectively.Figures 6 and 7 also include the longitudinal straintransverse strain relationship corresponding to the stress-strain relationships of figure 5.Figure 8 shows the typical change in shape of the transverse deformation profile of the top surface of the beams with load increasing to failure and figure 9 provides a schematic representation of the radial forces and stresses developing with increasing load due to the deflected shape of the beams. Typical load-deflexion relationships of the beams are shown in figure 10whereas figure 11 depicts the variation on critical sections of the average vertical strains measured on the side surfaces of the beams with the transverse strains measured on the top surface. Figure 12 indicates the strength and deformation response of a typical concrete under various states of triaxial stress and figure 13 presents the typical crack pattern of the beams at the moment of collapse. Finallyfigure 14 shows the shape of the longitudinal stress distribution on the compressive zone of a critical section at failure predicted on the basis of the concepts discussed in the following section.中文翻译I：在多向应力作用下从混凝土的特性看受弯钢筋混凝土梁变化的一个基本试验M. D. Kotsovos 伦敦皇家科学与技术学院土木工程系本文所探讨的问题是通常认为在荷载递增下钢筋混凝土结构呈现弹性状态这必须是因为混凝土的应力-应变关系有一个逐渐递减的临界部分的真实性试验数据显示受弯钢筋混凝土梁会在受压面的纵向压应变超出0.0035这表明这些应变是钢筋混凝土结构的本质它是由于一个比极限强度小的复杂多向的应力状态而不是塑性材料的特性引起的一个复杂应力系统的存在为梁的状态提供了一个基本试验而不是想象的一个现有设计过程1.引言"剖面"理论不仅是通常认为能很真实地描述钢筋混凝土梁和预应力混凝土梁在弯矩和轴向荷载下的变形而且能确切地阐述所以它提供了一个设计工具因为它的有效和简单而闻名[1]假设在临界横截面伤是均衡的这个理论分析地描述了一个梁的承载能力和几何特性之间的关系变形协调必须满足"水平横截面荏苒水平"的假定和纵向混凝土和钢筋的应力是通过材料的应力-应变的特性来估算的为了简化计算忽略横向的应力和应变受压混凝土的应力-应变特性认为能够被混凝土试块的变形充分地描述例如在极限的有限状态下棱柱体或圆柱体在横截面的受压区受单轴压力和应力就像现行规范所建议的CP110[1]显示出一个与图1相似的形状图1表明纵向应力随着与中和轴的距离增加而增加至最大值然后保持不变这个分布图已经达到安全性和受压混凝土的应力-应变关系的广泛观点由上升和逐渐下降的两部分组成(如图2所示)超出极限的部分材料的塑性应力能力如图1所示被认为对梁的最大承载能力有较大的作用图1.临界面破坏建议CP为110的应力和应变分布图2.受压混凝土结构的标准应力-应变关系然而最近关于在集中力作用下的混凝土的变化的一个分析性调查表明在压应力作用下混凝土的极限强度变形没有对所有被调查的结果形式的变化产生明显的影响([2][3])如果这个变化对任何结果都是典型的那么在钢筋混凝土梁的顶面被测的很大的压应变(超出量0.0035)在它的极限有限状态下(如图1)不能对极限单轴应力-应变特性产生作用因此因为压应变在单轴压力下的任何混凝土的极限强度等级下为ε=0.002(如图2所示)在混凝土的单轴应力-应变关系下降部分将出现一个在荷载作用下梁变化的现在可行的预测根据以上的观点本文的描述都在以下的评价中广泛的支持观点的一个单轴应力-应变关系由一个上升的和一个逐渐下降的部分组成对受弯的根据混凝土梁的变化的真实描述是非常必要的这个结果是从梁在两点荷载作用下弯曲得到表明很大的应变的通过梁受压的混凝土呈现的由于三维应力而不是一味的混凝土极限应力-应变特性这表明材料本身受到一个完整和直接的承载能力损失当极限强度被超过的假定与弹性结构的变化并存的通过偏心荷载或瞬间旋转关系表明的2.试验细节2.1试块三根矩形钢筋混凝土梁跨度915mm横截面为102mm51mm受剪区跨度为305mm(如图2所示)受力筋由两个直径为6mm屈服荷载为11.8kN的钢筋组成在梁端部钢筋弯起就能为整个受剪跨度提供抗力整个受剪跨度内压缩张拉的加强筋布置了七个直径为3.2mm的箍筋在梁的中间部分没有压缩加强筋和箍筋根据上面所述的钢筋布置所有的梁都是受弯破坏而不是受剪破坏尽管剪跨比为3所有的梁与受控的试块一起放在20 的湿麻袋下七天然后贮存在实验室条件下(2040%湿度)2个月直到试验结束所有混凝土配料都在表格I中2.2试验过程通过液压锤和分布梁加载每次大约增加0.5kN为了测量荷载和试块的形变每次持荷约2分钟荷载用一个荷载单元来测量形变由20mm长的电阻应变片和位移转换器测得应变片贴在梁纵向和横向的顶面和侧面(如图4所示)图4也表明了直流电压位移转换器（LVDT'S）的位置它是用来测量跨中和加载横截面的形变测量数据记录在计算机自动数据记录仪中能够测量应变和形变的灵敏度分别为±2微应变和±0.002mm3.试验结果主要的试验结果是从试验中得到的能更好地了解梁的变化所示图5 至图14的信息是必不可少的图5表明结果的单轴压应力-应变关系应用于调查中而图6 和图7表明纵向应变与横向应变的关系分别位于(a)弯曲裂缝最终导致破坏横截面出和(b)受剪区跨内的横截面出图6和图7也包含了纵向应变-横向应变与图5的应力-应变关系是一致的图8中标准的改变在梁顶面的横向形变轮廓图中和图9提供一个轴力和应力随着荷载的增加而增大导致梁向下变形的图框表示方法梁的标准偏心荷载关系如图10所示而图11描述了测得平均竖向应变的梁侧面的临界截面变形和横向应变在顶面测得图12中标准结果的强度和形变在各种状态的十三轴应力下河图13所呈现的梁标准裂缝图样在破坏的瞬间最后图14表明在临界截面的受压区伤纵向应力的分布形状可根据概念来预测破坏在以下部分将被讨论图3.梁的细节外文原文II:Some questions on the corrosion of steel in concrete.Part Ⅱ: Corrosion mechanism and monitoringservicelife prediction and protection methodsJ.A. GonzdlezS. FelifdP. RodffguezW. LfpezE. RamlrezC. AlonsoC.AndradeABSTRACTThis second part addresses some important issues that remain controversial despite the vast amounts of work devoted to investigating corrosion in concrete-embedded steel. Specificallythese refer to: 1) the relative significance of galvanic macrocouples and corrosion microcells in reinforced concrete structures; 2) the mechanism by which reinforcements corrode in an active state; 3) the best protective methods for preventing or stopping reinforcement corrosion; 4) the possibility of a reliable prediction of the service life of a reinforced concrete structure ; and 5) the best corrosion measurement and control methods. The responses provided are supported by experimental resultsmost of which were obtained by the authors themselves.1. INTRODUCTIONConcrete-embedded steel is known to remain in apassive state under normal conditions as a result of the highly alkaline pH of concrete. The passivity of reinforcements ensures unlimited durability of reinforced concrete (1KC) structures. Howeverthere are some exceptional conditions that disrupt steel passivity and cause reinforcements to be corroded in an active state. This has raised controversial interpretationssome of which were discussed in Part I of this series [1]. This Part II analyses though far from exhaustivelyother - to the authors minds at least - equally interesting issues on which no general consensus has been reached.2. MATERIALS AND METHODSThe reader is referred to Part I for a detailed description of the materials and methods used in this work. Most of the experimental results discussed herein were obtained with the same types of specimens and slabs.Galvanic couples were determined on speciallydesigned specimenssuch as those shown in Figs. 1 and 2.Near-real conditions were simulated by using a beam that was 160cm long and 7 x 10 cm in cross-section. The beam was made from 350 kg cement/m 3half of whichcontained no additiveswhile the other half included 3% CaC12 by cement weight [2](Fig. 1). In order to study the effect of the Sanod/Scathoa ratio on galvanic macrocouplesthey were modelled by surrounding a small carbon steel anode with a stainless steel (AISI 304) cathode and vice versa(Fig. 2). In this waythe ratio's consistensy was assured. In additionthe potential and icorr of stainless steal and those of the passive structures were very similar.Fig. 1 - Beam used to measure icoTr and Ecorr in Fig. 2 - Scheme of galvanic macrocouples embeddedconcrete with and without chlorides and to in chloride- containing mortar used to study theillustrate the significance of passive steel/active effect of the Sanod/Scathod ratio and their relativesteel macrocouples. significance to corrosion microcells.3. RESULTS AND DISCUSSION3.1 What is the relative significance of galvanic macrocouples and corrosion microcells in RC structures ?According to several authors [35]the polarization resistance method provides an effective means for estimating the corrosion rate of steel in PC ; the method is quite rapidconvenientnon-destructivequantitative and reasonably precise. Howeverit is uncertain whether it may give rise to serious errors with highly-polarized electrodes by the effect of passive/active area galvanicmacrocouples in the reinforcements [6].Based on the authors' own experience with the behaviour of galvanic macrocouples in PCthe contribution of these macrocouples to overall corrosion is very modest rehtive to that of the corrosion microcells formed in the active areas of reinforcements in the presence of sufficient oxygen and moisture [278]. Thusit has been experimentally checked that:(a) Galvanic macrocouples have a slight polarizing effect on anodic areas in wet concretewhose potential is thereby influenced in only a few millivolts.(b) On the other handmacrocouples have a strong polarizing effect on passive areas despite the low galvanic currents involved relative to the overall corrosion current.(c) As a resultgalvanic currents can result in grossly underestimated icorr values for the active areas since they are often smaller than 10% of the ico= values estimated from polarization resistance measurements.(d) The corrosive effect ofcoplanar macrocouples on RC structures only proves dangerous within a small distance from the boundary of active and passive areas. Fig. 3 compares the estimated icorr and ig valuesin mortar containing 3 o~ A CaC12per anode surface unit for a number of anode/cathode surface ratios for AISI 304 stainless steel/carbon steel macrocouples in support of the above conclusions [9].3.2 By what mechanism do reinforcements corrode in an active state ?When the passive state is lostthe rate of reinforcement corrosion in inversely proportional to the resistivity of concrete over a wide resistivity range [10]. BecauseFig. 3 - Relative significance of corrosion microcells Fig. 4 - Trends in ico. and Ecorr for(icorr) and galvanic macrocouples (i.) in corrosion specimens exposed to an oxygen-freeof steel embedded in mortar containing no chloride. environment.Both currents were calculated relative to Sanod(carbon steel in the macrocouples of Fig. 2).the environment's relative humidity and ionic additives of concrete determine concrete resistivitythese factorstogether with oxygen availability at reinforcement surfacescontrol the corrosion rate [11].The electric resistivity of water-saturated concrete structures is relatively very lowand the corrosion rate is believed to be essentially controlled by the diffusion of dissolved oxygen through the concrete cover up to reinforcements. This is consistent with the widespread belief that the sole possible cathodic reaction in neutral and alkaline solutions is oxygen reduction.The significance ascribed to the role of oxygen justifies the efforts to determine its diffusion coefficient in concrete[1213]. The variety of methods and experimental conditions used for this purpose have led to a wide range of diffusivity values (from 10 -12 to 10 -8 m2/s) for oxygen in cement paste [14].Since the diffusion coefficient of oxygen in aqueous solutions (1)O2 = 10 -5 cm2/s-1)is saturation concentration (CO2 = 2.1 x 10 -7 mol/cm 3) and the approximate thickness of diffusion layers in stagnant solutions (8 = 0.01 cm) are wellknownthe limiting diffusion current can be calculated as :ilo2 = - z FD02C02/r = 8 x 10 -4 A/cm 2 (80 pA/cm 2)where z is the number of equivalents per mole (4) and F the Faraday (96500 A.s/eq).For 1-cm thick mortar covers of average porosity 15%(see Fig. 1 in Part I) [1] and a diffusioja layer thickness of the same order as the cover thickness11o2 = 0.12 laA/cm 2which is quite consistent with the icorr values estimated under pore saturation conditions at the end of the curingprocessboth for mortars containing no chloride ions and for those including 24 or 6% C1- [16].On the other handicorr values of ca. 10 liA/cm 2 (see Fig. 9 in Part I) [4] have been obtained by several authors for mortars with chlorides or carbonated mortars which are incompatible with the rates allowed by the limiting diffusion current of oxygen. Thereforein some circumstancesalternative cathodic processes allowing for faster kinetics must therefore be involved. In recent workthe concurrence of creviceschloride ions and dissolved oxygen at the steel/concrete interface was claimed to provide the thermodynamic conditions required for protons to be reduced and the alternative mechanism to occur [1117].There are a number of facts that refute oxygen reduction as being the sole corrosion rate-determining stepnamely:- Under some circumstancesonce corrosion in an activestate has startedit develops at the same rate even though oxygen is being removed from the medium (Fig. 4) [11].- As saturation of concrete pores decreaseconcrete resistivity controls ico~r over a wide resistivity range ; therefore the corrosion rate seems to decrease in proportion to the ease with which oxygen penetrates into the structure(Fig. 5)[10].On the other handthere are several arguments in favour of proton reduction in Ca(OH)2-saturated solutions or cement mortars [11] :- The pH decreases from 12.6 to ca. 5 within crevices at the steel/electrolyte interface upon exposure of the steel to a Ca(OH)2-saturated solution with C1- additions and wellaerated. If sufficient oxygen is availablethe pH can drop as low as 1-2.- The emergence of acid exudates ofpH 1-5 from cracks and macropores in chloride-containing mortar specimens under wet atmospheres at high corrosion rates (5-10 pA/cm2).- The formation of gas bubbles over iron hydroxide membrane-coated pits when the steal is polarized anodically in a Ca(OH)2-saturatedchloride-contaminated solution at potentials below those required for oxygen release. Everything points to pits with a low enough pH for the anodic current applied to overlap with a corrosion process involving proton reduction as a cathodic half-reaction.When concrete-embedded steel is corroded in an active stateits corrosion kinetics rise exponentially with increasing pore saturation (Fig. 6) similarly to atmospheric corrosion in bare steel as the environment's relative humidity increases [18]. At some points in the reinfor- cementsa catalytic cycle may take placee.g.those put forward by Schikorr for atmospheric corrosion of steel [19]with chloride ion rather than SO2-as the catalyst (Fig. 6).Fig. 5 - Relationship between mortar resistivity Fig. 6 - Influence of the degree of pore saturationand the corrosion rate of reinforcements. on the corrosion rate of reinforcements.中文翻译II:混凝土中钢腐蚀的有关问题Ⅱ：腐蚀机理和监督、使用年限的预测和保护方法J.A. GonzdlezS. FelifdP. RodffguezW. LfpezE. RamlrezC. AlonsoC.Andrade摘要:第二部分阐述几个仍然存在争议的重要问题尽管已经在混凝土中钢腐蚀的调查研究投入了大量的工作特别是这几方面：1)在钢筋混凝土结构中的大电偶和腐蚀微电池对的相对重要性；2)激活状态的钢筋腐蚀机理；3)阻止或停止钢筋腐蚀最好的保护方法；4)一个钢筋混凝土结构使用年限的可靠预测的可能性探索；5)最好的防腐措施和控制方法这些回答需要试验得出大部分都由作者们得出1.前言正常条件下强碱混凝土中的钢仍然处于钝化状态钢筋的钝性能保证钢筋混凝土结构无限的耐久性然而有一些能破坏钢的钝性和引起钢筋腐蚀的实验条件在第Ⅰ部分中讨论到的一些实验结构已经引起了很多争论[1]第Ⅱ部分的分析虽然没有竭尽全力但至少是作者的意思就像有趣的问题有不同的意见一样2.材料和方法读者指出在第Ⅰ部分详细描述了用于这项工作的材料和方法这里所讨论的大部分实验结果都是从一样的试块和平板中得到的电偶是由特殊设计的试块确定的如图1和2所示用一根长16m70mm×100 mm横截面的梁模拟近真实条件梁是由每立方米350kg水泥制成梁的一半含有添加剂另一半含有水泥的重量的3%的CaCl2[2](图1)为了了解S正极/S负极的比值对大电偶的影响用在一个小的碳素钢正极环绕一个不锈钢负极并夹紧来模拟这样比值的连贯性是可靠的此外与钝化结果的电位和不锈钢的icorr是非常相似的图1.梁用来分别测量混凝土中含有和不含有氯化物图2.用电耦合牢牢嵌入含有氯化物的砂浆里来研究的icorr和Ecorr来说明钝化钢/活跃钢耦合的意义S正极/S负极的作用和腐蚀微电池对的相对意义的方案3.结果和讨论3.1什么是在钢筋混凝土结构中大电偶和腐蚀微电池对的相对重要性？根据一些作者[35]极化电阻作用为估计钢筋混凝土中腐蚀速度提供了一个有效的方法；这个方法是非常快、方便、非破坏性、适量和相当精确的然而它不确定是否会对高度极化的电极产生严重的错误通过在钢筋中的大电偶的钝化面积与激活面积的比值的影响在作者自己对钢筋混凝土中大电偶性质的实验基础上这些大电偶对所有的腐蚀是非常适度的与存在充分的氧气和水分条件下腐蚀微电池对形成激活状态的钢筋比较[278]因此它已被实验验证：(a)大电偶对潮湿混凝土中的阳极部分由一个轻微的极化作用只要几毫伏就可以影响它的电位(b)在另一方面大电偶对钝化部分有一个很强的极化作用尽管低电流的运用相对于所有腐蚀流(c)因此电流可能会导致非常低估在激活部分的icorr的值因为它们通常比极化电阻值估算的icorr值的10%还小(d)腐蚀剂会引起钢筋混凝土结构上共面的电偶只能证明从激活面积到钝化面积边缘的一个很短的距离存在危险图3是估算的icorr与ig值的比较在砂浆中含有3%的CaCl2每个正极表面单元体为许多正极/负极表面比值作为美国钢铁学会304不锈钢/碳素钢电偶的一部分支持以上结论图3.腐蚀微电池对(icorr)和电耦合(ig)在包裹在图 4.暴露在自由氧环境下试块的icorr和Ecorr不含有氯化物砂浆里的钢腐蚀中的相对意义的变化趋势电流都是相对于S负极而计算得到的(在图2的电耦合中的碳素钢)3.2钢筋腐蚀的机理是什么？当钝化状态消失钢筋的腐蚀速度与混凝土的电阻率成反比例在一个很宽的电阻率范围内[10]因为环境中的相对湿度和混凝土的离子型外加剂确定混凝土的电阻率这些因素与氧气一起在钢筋的表面控制着腐蚀速度[11]饱和水混凝土结构的电阻率是相对非常低的而且腐蚀速度实际上是溶解氧的扩散控制的通过混凝土包住钢筋实现这与在中性和强碱条件下唯一可能的负极反应是氧气的还原作业这个理念是一致的这个重要性归因于氧气的循环作业它证明这些作用对确定它在混凝土中的扩散率是正确的[1213]各种方法和实验条件用于这个目的已得出了一定范围的水泥浆中的氧气的扩散率(从10-12到10-8m2/s)[14]因为水溶液(CO2=10-5cm2/s-1)中氧气的扩散率是饱和浓度(CO2=2.1×10-7mol/cm3) 不流动环境中(?=0.001cm)扩散层的近似密度都是众所周知的这个有限扩散流可以这样计算：其中z是等价的每摩尔(4)的数值而F就是法拉第(96500A?s/eq)平均孔隙率为15%的1cm厚的砂浆保护层厚度与扩散层厚度一样与在养护期的最后空隙饱和条件下估算得的icorr值是非常一致的这些砂浆不含氯化物离子而都含有24或6%的Cl-[16]另一方面ca.10?A/cm2的icorr(见第Ⅰ部分图9)[4]已经由一些作者从含氯化物的砂浆或碳酸盐砂浆与氧气有限的扩散流所允许的速度是不协调的因此在一些环境下替代负极的过程必须有更快的动力在最近的工作中裂缝、氯化物例子和溶解氧并存在钢与混凝土的交界面可以为质子的还原和替换机理的发生提供热动力条件[1117]有很多论据反驳氧气的还原作用作为底面腐蚀的定速步骤即：- 在一些环境下腐蚀一旦开始它发展到同一个速度尽管氧气正在从媒介中排除(图4)[11]- 当混凝土空隙饱和作用降低混凝土的电阻率控制icorr在一个宽泛的电阻率范围内；因此腐蚀速度的减小好像与氧气进入结构的难易成反比例(图5)[10]在另一方面有一些论点支持在饱和Ca(OH)2中或水泥砂浆中的质子还原反应[11]：- PH值由12.6减小到ca.5在暴露的含有Cl-的饱和Ca(OH)2中的钢与电解质溶液的交界面上如果提供充足的氧气PH值可以降低到1-2- 从在潮湿的空气中含有氯化物的砂浆试块的裂缝和大空隙中暴露的PH值1-5的酸性分泌物腐蚀速度很快(5-10?A/cm2)- 在蚀坑处涂上氢氧化铁膜的钢在含有氯化物的饱和Ca(OH)2中极化成阳极时会产生气泡因为电位的降低需要释放氧气每一个蚀坑点有一个足够低的PH因参与质子还原反应就像阴极半反应它们的腐蚀过程与阳极流互相重叠当包裹在混凝土中的钢处于腐蚀状态它的腐蚀动力指数随着空隙饱和作用的上升而升高(图6)就像裸露在大气中的钢的腐蚀随着环境的相对湿度的上升而增加一样[18]在钢筋上的一些点催化循环可能被取代等这些是由Schikorr提出的钢的大气腐蚀[19]是氯化铁而不是SO42-作为催化剂(图6)图5.砂浆电阻与钢筋腐蚀速度的相互关系图6.孔隙饱和度对钢筋腐蚀速度的影响????????宁波工程学院毕业设计（论文）1。

土木工程--毕业设计外文翻译(原文+翻译)

毕业设计（论文）外文翻译题目西北物流中心2号楼设计专业土木工程班级土木074学生指导教师二零一零年Low-coherence deformation sensors for themonitoring of civil-engineering structuresD. Inaudi a, A. Elamari b, L. Pflug a, N. Gisin b, J. Breguet b, S. Vurpillot a “IMAC, Laboratory of Stress Analysis, Swiss Federal Institute of Technology, CH-1015 Lausanne, Switzerland ‘GAP, Group of Applied Physics -Optical Seciion, Geneva University CH-1205 Geneva, SwitzerlandRcccivcd 25 January 1993; in revised form 8 March 1994; accepted 25 March 1994 AbstractAn optical-fiber deformation sensor with a resolution of 10 pm and an operational range of 60 mm has been realized. The system is based on low-coherence interferometry instandard single-mode telecommunication fibers. It allows the monitoring of large structures over several months without noticeable drift. No continuous measurement is needed and the system is insensitive to variations of the fiber losses. This technique has been applied to the monitoring of a 20 m X5 m X0.5 m, 120 ton concrete slab over six months. It is possible to measure the shrinkage of concrete and its elastic coefficient during pre-straining, giving reproducible results in good agreement with theoretical calculations and measurements performed on small concrete samples. This paper describes the optical arrangement and the procedures used to install optical fibers in concrete.Keywor&: Ikformation sensors; Civil-engineering structures1. IntroductionBoth the security of civil-engineering works and the law require a periodic monitoring of structures. The methods used for this purpose, such as triangulation, water levels or vibrating strings, are often of tedious application and require one or many specialized operators. This complexity and the resulting costs limit the frequency of the measurements. Furthermore, the spatial resolution is often poor and the observation is usually restricted to the surface of the object. There is thus a real demand for a tool allowing an internal, automatic and permanent monitoring of structures with high accuracy and stability over periods typically of the order of 100 years for bridges. In this framework, fiber-optic smart structures (i.e., structures with self-testing capabilities) are gaining in importance in many fields including aeronautics and composite material monitoring. This technology can be applied in civilengineering and in particular for the short- and long-time observation of large structures such as bridges, tall building frames, dams, tunnels, roads, airport runways, domes, pre-stressing and anchorage cables. The monitoring of such structures requires the development of a measuring technique with high accuracy,stability and reliability over long periods. It has to beindependent of variations in the fiber losses and adapted to the adverse environment of a building site. To reduce the cost of the instrumentation, it is furthermore desirable to use the same portable reading unit for the monitoring of multiple structures. We describe here asystem based on low-coherence interferometry responding to all these requirements.2. Experimental arrangementThe measuring technique relies on an array of standard telecommunication optical fibers in mechanical contact with concrete. Any deformation of the host structure results jn a change in the optical length of he fibers. Each sensor line consists of two single-mode ibers: one measurement fiber in mechanical contact with the structure (glued or cemented) and a reference iber placed loose near the first one (in a pipe) in order to be at the same temperature. Since the measurement technique monitors the length difference beween these two fibers, only the mechanical deformation will have an effect on the results while all other perurbations, such as thermally induced changes in the refractive index of the fibers,will affect the two in an identical way and cancel each another out. To measure the optical path difference between the two fibers, a low-coherence double interferometer in tandem configuration has been used (Fig. 1) [l]. The source is an LED (light-emitting diode) working around 1.3 pm with a coherence length L, of 30 pm and a rated power of 200 pW. The radiation is launched into a single-mode fiber and then directed toward the measurement and the reference fibers by means of a 50:50 single-mode directional coupler. At the ends of the fibers two mirrors reflect the light back to the coupler, where the beams arc recombined with a relative delay due to the length difference AL, between the fibers, and then directed towards the second (reference) interferometer. The reference interferometer is of Michelson type with one of the arms ended by a mobile mirror mounted on a micromctric displacement table with a resolution of 0.1 pm and an operating range of 50 mm. It allows the introduction of an exactly known path difFcrence AL, between its two arms. This fiber interferometer is portable and needs no optical adjustment after transportation. It has been developed by the GAP with the support of the Swiss PTT for optical cable testing [2].The intensity at the output of the reference inter- ferometer is measured with a pig-tail photodiode and is then given by [3]where zz,,r is the effective refractive index of the fiber, zzg the group refractive index (about 1% higher than nefr in silica), A, the central vacuum wavelength of the light, zi,, the autocorrelation function taking the spectral characteristics of the emission into account and AL the physical path difference between the two interfering paths. Further similar interference terms appear in Eq.(1) in the special cases when AL, <L, or AL, < L,. When the optical path difference between the arms in the reference interferometer corresponds to the one induced by the two fibers installed in the structure (within the coherence length of the source), interference fringes appear. Scanning AL, with the mirror of the reference interferometer it is possible to obtain AL = 0either with AL, = AL, or with AL, = -AL,, and thus two interference fringe packets as described by Eq. (1). The mirror position corresponding to AL, = 0 also produces an interference and is used as a reference. These three fringe packets arc detected by means of a lock-in amplifier synchronized with the mirror displacements. The mirror displacements and the digitalization of the lock-in output are carried out by means of a portable personal computer. Since the reference signal is gcnerated separately and does not have a constant phase relation to the interference signal, only the envelope of the demodulated signal has a physical meaning and corresponds to the envelope of the fringe pattern. A lock-in plot showing the three typical peaks is shown in Fig. 2. Each peak has a width of about 30 pm. The calculation of its center of gravity determines its position with a precision better than 10 pm. This precision is the limiting factor of the whole measurement technique. Since AL, is known with micrometer precision, it is possible to follow AL, with the same precision.Fig. 1. Experimental setup of the low-coherence double Michelson interferomctcr. D. Innudi et al. 1 Semors andFig. 2. Typical fringe cnvclope as a function of the mirror position. The distance between the central and the lateral peaks corresponds to the length difference between the measurement and the reference fibers mounted in the table. Any change in the length of the structure results in a change in the position of these peaks. Any change in the losses of the fibers will result in a change of the height of the peaks. The central peak is fixed and used as a reference.The path difference AL, is proportional to the de-formation of the structure AL, with the relation between the two given by [4]where p is Poisson’s ratio and pij is the strain optic tensor (Pockcl’s coefhcients). The coefficient 5 takes into account the variation of the effective index neff in a fiber under strain.A degradation of one or both fibers (due to aging, for example) will result in a lower visibilityof the fringes but will not affect its position. The information about the deformation of the structure is encoded in the coherence properties of light and not in its intensity as in the majority of the sensors applied to date in civil-engineering structures, mostly based on microbend losses and/or optical time-domain reflectometry (OTDR) techniques. Interference peaks resulting from reflections as low as -30 dB of the source power can be detected by our system without phase modulators. By modulating the phase in one of the four arms of the two interferometers, one can increase the dynamic range of the device to more than 100 dB [5].Even if the polarization dispersion and bend-induced birefringence in the sensing fibers could reduce the visibility of the interference fringes or even split the fringe packets, none of those effects was observed in our experiment. No adjustment of polarization between the reference and the sensing arm was then necessary. A good mechanical contact between the measurement fiber and the structure under test is fundamental. In this study a number of installation procedures have been tested and optimized for the different measurements (shrinkage, elasticity modulus, etc.). The mounting techniques can be divided into two main categories: full-length coupling and local coupling.During our tests five out of six optical fiber pairs with a 0.9 mm nylon coating, being mounted on the external face of a 20 m long plastic pipe and protected only with thin rubber bands (see Fig. 3(a)), survived the concreting process. During the setting process the concrete envelops the fiber and realizes the desired mechanical contact. Those fibers showed a minor increase in the scattering losses and the appearance of small parasite peaks. The measurements on those fibers were consistent with the results obtained with other installation techniques (see below). It seems that for full-length coupling the nylon coating transmits the structure deformations (extension and shortening) entirely to the fiber core. This installation technique is very promising when compared to the usual procedure, consisting of a pipe protecting the fibers during the pouring of concrete and being removed before the setting process begins. This second method seemsmore adapted to small samples than to full-scale structures. Eleven otherfiber pairs were glued at the two ends of the table after removing locally the protective coating layers of the fibers (see Fig. 3(b)). The silica fiber was ftxed with epoxy glue to a metallic plate mounted on the end facesof the concrete structure. The gluing length was about 20 mm. Apre-strain (between 0.1 and 0.4%) has been given to those fibers during the gluing process to keep them under tension and allow the measurement of both expansion and shrinkage of the structure. This type of local coupling proved to be the most reliable, but was not adapted to following thedeformation during the pre-stressing of the table because of the important surface deformations occurring during this operation. The problem has been overcome by gluing other fibers inside the pipes at about two meters from the surfaces, i.e., far from the force insertion region (see Fig. 3(c)).Fig. 3. Schematic representation of three of the installation techniques used:(a) direct concreting of the measurement fiber mounted on a plastic pipe; (b) fiber glued at the table surface; (c) fiber glued inside the pipe at 2m from the pipe ends.Fig. 4. Top and side views of the concrete table measured in the experiment and position of the sensing-fiber pairs A, B, C and D. Fibers A, B and C arc glued at the surface of the structure, while fiber D is glued inside a pipe, 2 m away from the surface of the slab. Twelve more fihcr pairs were installed, but are not shown for simplicity.To study the possible effect of creep in strained fibers [6], one fiber has been mounted on a mechanical support that allows the fiber to be tightened only at the time of the measurement. No difference between this fiberand those permanently strained has been observed over a period of six months, confirming the assumption that no creep occurs for fiber strains below 1%. Since the scanning range of the mirror is 5 mm, it was easy to cleave the 20 m long fibers within this margin. The Fresnel reflection of the cleaved fibers combined with the high dynamic of the system allow a measurement of AL,,. This value of AL, can than be used to correct the cutting and obtain pairs with length differences below 1 mm. Two ferrules were then installed on the fiber ends and mounted in front of a polished inox surface. Chemical silver deposition was also used to produce mirrors on the cleaved fiber ends.Fig. 6. Comparison between the measurements performed on the structure by optical fibers and the ones performed on 360 mm and 500 mm samples in a mechanical micrometer comparator. The measurement on the samples was possible only during the first two months.3. ResultsSeveral long- and short-term measurements have been carried on a 20 m x 5 m x 0.5 m, 120 ton concrete slab intended to be used as a vibration-isolated base for optical analysis (in particular by holographic and speckle interferometry) of large structures [7].This structure has been concreted indoors, allowing controlled environmcntal conditions and known concrete composition to be achieved. Samples have been prepared with the same material composition and are under permanent test for their mechanical properties (resistance, shrinkage and elastic coefficient). This allows a direct comparison between the results on the full-scale structure and the samples. The table has been pre-strained 23 days after concreting in both length and width. It was possible at this time to measure the elastic coefficient of the material in full scale. Fig. 4 shows a schematic representation of the table and the position of the fibers referred to in the experimental results. At the time of writing, the table has been under test for six months. Over this period the shrinkage in the longitudinal direction (i.e., over 20 m) has been about 6 mm. We show in Fig. 5 the results of the measurements for three (glued) fibers over 175 days. The table has a T profile (Fig. 4). It is evident from Fig. 5 that thefibers mounted near the borders of the table, i.e., were the thickness is smaller, registered a larger shrinkage, as expected according to the concrete theory. Adjacentfibers give consistent results independently of the installation technique. No difference has been noticed between the fibers under permanent tension and those loosened between the measurements, suggesting that no creep of glass fibers occurred. The shrinkage measured with the fiber system has been compared during the first two months with the results obtained with a mechanical comparator mounted on two samples of 360 mm and 500 mm, respectively.The observed deformations have been scaled to 20m and are compared in Fig.6 to the results obtained with fibers B and C. Very good agreement is found between the two measurements. A theoretical comparison between the experimentalresults and the Swiss civil engineering standards has also been carried out. The experimental data and the standards are in agreement within f 10%. A more accurate simulation including the physico-chemical properties of the concrete used is under development. The table was pre-stressed 23 days after concreting. The five steel cables running over the length of the table and the forty cables running over its width were stretched with a force of 185 kN (18.5 Tons) each. The fibers glued to the surface and those in direct contact with concrete over the whole length measured an expansion of the table instead of the expected shrinkage. This is due to the important surface deformations occurring near the force-insertion points, i.e., near the pre-stress heads that were placed near the fiber ends. Fiber D glued inside the plastic pipe at 2m from each endwas not subject to these local effects and measured a shortening of 0.23 mm. The theoretical calculation based on an elastic coefficient of 30 kN/mm2gives a shortening of 0.28mm at the borders and 0.19 mm at the center of the table. Since fiber D was placed in an intermediate position, the experimental value can be considered to be in good agreement with the theory.4. ConclusionsA new deformation sensor adapted to the monitoring of civil-engineering structures has been proposed. it is based on low-coherence interferometry in standard lowcost telecommunication fibers. The resolution of the measurements is 10 pm, the operational range is 60mm and the stability has been tested over six months without noticeable drift. The reading unit is compact and portable, needing no optical alignment before the measurements. It is controlled by a portable personal computer, which is also responsible for the data trcatment. The same reading unit can be used to monitor multiple fiber lines by simple manual unplugging. This technique is furthermore practically insensitive to increased losses due to degradation of the fibers. A test study has been carried out on a 20m ~5m X 0.5m concrete slab, giving consistent results when compared to other measurement techniques based on samples or to concrete theories. It was possible to follow concrete shrinkage over six months (the cxper- iment will continue for about five years) and to measure the elastic coefficient on the full-scale structure. Different fiber-installation techniques adapted to the measurement of various parameters have been tested in building-site conditions. This technique appears very promising for the mon-itoring of civil-engineering structures such as bridges, dams and tunnels, allowing internal, automatic and permanent monitoring with high precision and stability over long periods.AcknowledgmentsThe authors are indebted to R. Passy and R. Delez for their assistance, encouragement and helpful dis-cussion. We acknowledge the IMM Institute in Lugano (Switzerland) for placing the table at our disposal and for the measurements carried out on concrete samples. We are grateful to Dr M. Pedretti and Ing R. Passera for their personal engagement in the project. We also thank Cabloptic in Cortaillod (Switzerland) for sup-plying all the optical fibers used in the experiment. This research has been performed with the financial support of CERS (Commission pour 1’Encouragement de la Recherche Scientifique).References[1] A.Koch and R.Ulrich,Fiber optic displacement sensor with 0.02mm resolutionbuy white-light interferometry,sensors and actuators A,25-27(1991)201-207[2]N.Gisin,J.-P.Von der weid and J.-P.Pellaux,Polarization mode dispersion ofshort and long single-mode fibers,J.Lightwave technol,9(1991)821-827.[3] A.S.Gergcs,F.Farahi,T.P.Newson,J.D.C.Jones and D.A.Jackson, Fiber-opticinterferometric sensors using low coherence source:dynamic range enhancement,Int. J.Op-toelectron,3(1988)311-322.[4] C.D.Butter and G.B.Hacker, Fiber optics strain gauge,Appl.Opt,17(1978)2867-2869.[5]H.H.Gilger,G.Bodmer and Ch.Zimmer, Optical coherance domain retlectometry asa test method of integrated optics devices,Proc.2nd Opt. Fibre Meas. Conf:OFMC 93, Turin, Ztuly, Z993, pp.143-146.[6]J.-P.Jaguin and A.Zaganiaris,La mecanique de rupture appliquee aux fibresoptiques, Verres Refract, 34 (Jul-Aout)(1980).[7]L.Pflug and M.Pedretti, Construction of a loo-tonnes holographictable,ZS&TISPIE Znt.Symp. Electronic Imaging, SanJose,CA,USA,1993,pp.50-54.传感器和执行器 A 44（1994）12.5-130用低变形传感器监测民用工程结构变形的一致性D.Inaudi a, A.Elamari b, L.Pflug b, N.Gisin b, J.Breguet b, S.Vurpillot aa IMAC、实验室的应力分析,瑞士联邦理工学院,CH-1015瑞士洛桑b GAP,群应用物理-光学部分,日内瓦大学，CH-1205瑞士日内瓦举行1993年1月25日实验;1994年3月8日修订,1994年3月25日发表文摘一个光纤变形的分辨率的传感器,10µm和运行范围的60毫米已经实现了。

土木工程毕业设计外文翻译

外文原文Talling Building And Steel Construction Although there have been many advancements in building construction technology in general. Spectacular archievements have been made in the design and construction of ultrahigh-rise buildings. The early development of high-rise buildings began with structural steel framing.Reinforced concrete and stressed-skin tube systems have since been economically and competitively used in a number of structures for both residential and commercial purposes.The high-rise buildings ranging from 50 to 110 stories that are being built all over the United States are the result of innovations and development of new structual systems.Greater height entails increased column and beam sizes to make buildings more rigid so that under wind load they will not sway beyond an acceptable limit.Excessive lateral sway may cause serious recurring damage to partitions,ceilings.and other architectural details. In addition,excessive sway may cause discomfort to the occupants of the building because their perception of such motion.Structural systems of reinforced concrete,as well as steel,take full advantage of inherent potential stiffness of the total building and therefore require additional stiffening to limit the sway.Systems in concrete. While tall buildings constructed of steel had an early start, development of tall buildings of reinforced concrete progressed at a fast enough rate to provide a competitive chanllenge to structural steel systems for both office and apartment buildings.Framed tube. As discussed above, the first framed tube concept for tall buildings was used for the 43-story DeWitt Chestnut Apartment Building. In this building ,exterior columns were spaced at 5.5ft (1.68m) centers, and interior columns were used as needed to support the 8-in . -thick (20-m) flat-plate concrete slabs.Tube in tube. Another system in reinforced concrete for officebuildings combines the traditional shear wall construction with an exterior framed tube. The system consists of an outer framed tube of very closely spaced columns and an interior rigid shear wall tube enclosing the central service area. The system (Fig .2), known as the tube-in-tube system , made it possible to design the world’s present tallest (714ft or 218m)lightweight concrete building ( the 52-story One Shell Plaza Building in Houston) for the unit price of a traditional shear wall structure of only 35 stories.Systems combining both concrete and steel have also been developed, an examle of which is the composite system developed by skidmore, Owings &Merril in which an exterior closely spaced framed tube in concrete envelops an interior steel framing, thereby combining the advantages of both reinforced concrete and structural steel systems. The 52-story One Shell Square Building in New Orleans is based on this system.In a steel structure,for example,the economy can be defined in terms of the total average quantity of steel per square foot of floor area of the building.Curve A in Fig .1 represents the average unit weight of a conventional frame with increasing numbers of stories. Curve B represents the average steel weight if the frame is protected from all lateral loads. The gap between the upper boundary and the lower boundary represents the premium for height for the traditional column-and-beam frame.Structural engineers have developed structural systems with a view to eliminating this premium. Systems in steel. Tall buildings in steel developed as a result of several types of structural innovations. The innovations have been applied to the construction of both office and apartment buildings.Frame with rigid belt trusses. In order to tie the exterior columns of a frame structure to the interior vertical trusses,a system of rigid belt trusses at mid-height and at the top of the building may be used.A good example of this system is the First Wisconsin Bank Building(1974) in Milwaukee.Framed tube. The maximum efficiency of the total structure of a tallbuilding, for both strength and stiffness,to resist wind load can be achieved only if all column element can be connected to each other in such a way that the entire building acts as a hollow tube or rigid box in projecting out of the ground. This particular structural system was probably used for the first time in the 43-story reinforced concrete DeWitt Chestnut Apartment Building in Chicago. The most significant use of this system is in the twin structural steel towers of the 110-story World Trade Center building in New York.Column-diagonal truss tube. The exterior columns of a building can be spaced reasonably far apart and yet be made to work together as a tube by connecting them with diagonal members interesting at the centre line of the columns and beams. This simple yet extremely efficient system was used for the first time on the John Hancock Centre in Chicago, using as much steel as is normally needed for a traditional 40-story building.Bundled tube. With the continuing need for larger and taller buildings, the framed tube or the column-diagonal truss tube may be used in a bundled form to create larger tube envelopes while maintaining high efficiency. The 110-story Sears Roebuck Headquarters Building in Chicago has nine tube, bundled at the base of the building in three rows. Some of these individual tubes terminate at different heights of the building, demonstrating the unlimited architectural possibilities of this latest structural concept. The Sears tower, at a height of 1450 ft(442m), is the world’s tallest building.Stressed-skin tube system. The tube structural system was developed for improving the resistance to lateral forces (wind and earthquake) and the control of drift (lateral building movement ) in high-rise building. The stressed-skin tube takes the tube system a step further. The development of the stressed-skin tube utilizes the façade of the building as a structural element which acts with the framed tube, thus providing an efficient way of resisting lateral loads in high-rise buildings, and resulting in cost-effective column-free interior space with a high ratioof net to gross floor area. Because of the contribution of the stressed-skin façade, the framed members of the tube require less mass, and are thus lighter and V less expensive. All the typical columns and spandrel beams are standard rolled shapes,minimizing the use and cost of special built-up members. The depth requirement for the perimeter spandrel beams is also reduced, and the need for upset beams above floors, which would encroach on valuable space, is minimized. The structural system has been used on the 54-story One Mellon Bank Center in Pittburgh.Steel construction refers to a broad range of building construction in which steel plays the leading role. Most steel construction consists of large-scale buildings or engineering works, with the steel generally in the form of beams, girders, bars, plates, and other members shaped through the hot-rolled process. Despite the increased use of other materials, steel construction remained a major outlet for the steel industries of the U.S, U.K, U.S.S.R, Japan, West German, France, and other steel producers in the 1970s.Early history. The history of steel construction begins paradoxically several decades before the introduction of the Bessemer and the Siemens-Martin (openj-hearth) processes made it possible to produce steel in quantities sufficient for structure use. Many of problems of steel construction were studied earlier in connection with iron construction, which began with the Coalbrookdale Bridge, built in cast iron over the Severn River in England in 1777. This and subsequent iron bridge work, in addition to the construction of steam boilers and iron ship hulls , spurred the development of techniques for fabricating, designing, and jioning. The advantages of iron over masonry lay in the much smaller amounts of material required. The truss form, based on the resistance of the triangle to deformation, long used in timber, was translated effectively into iron, with cast iron being used for compression members-i.e, those bearing the weight of direct loading-and wrought iron being used for tension members-i.e, those bearing the pull of suspendedloading.The technique for passing iron, heated to the plastic state, between rolls to form flat and rounded bars, was developed as early as 1800;by 1819 angle irons were rolled; and in 1849 the first I beams, 17.7 feet (5.4m) long , were fabricated as roof girders for a Paris railroad station.Two years later Joseph Paxton of England built the Crystal Palace for the London Exposition of 1851. He is said to have conceived the idea of cage construction-using relatively slender iron beams as a skeleton for the glass walls of a large, open structure. Resistance to wind forces in the Crystal palace was provided by diagonal iron rods. Two feature are particularly important in the history of metal construction; first, the use of latticed girder, which are small trusses, a form first developed in timber bridges and other structures and translated into metal by Paxton ; and second, the joining of wrought-iron tension members and cast-iron compression members by means of rivets inserted while hot.In 1853 the first metal floor beams were rolled for the Cooper Union Building in New York. In the light of the principal market demand for iron beams at the time, it is not surprising that the Cooper Union beams closely resembled railroad rails.The development of the Bessemer and Siemens-Martin processes in the 1850s and 1860s suddenly open the way to the use of steel for structural purpose. Stronger than iron in both tension and compression ,the newly available metal was seized on by imaginative engineers, notably by those involved in building the great number of heavy railroad bridges then in demand in Britain, Europe, and the U.S.A notable example was the Eads Bridge, also known as the St. Louis Bridge, in St. Louis (1867-1874), in which tubular steel ribs were used to form arches with a span of more than 500ft (152.5m). In Britain, the Firth of Forth cantilever bridge (1883-90) employed tubular struts, some 12 ft (3.66m) in diameter and 350 ft (107m) long. Such bridges and other structures were important in leading to the development and enforcementof standards and codification of permissible design stresses. The lack of adequate theoretical knowledge, and even of an adequate basis for theoretical studies, limited the value of stress analysis during the early years of the 20th century,as iccasionally failures,such as that of a cantilever bridge in Quebec in 1907,revealed.But failures were rare in the metal-skeleton office buildings;the simplicity of their design proved highly practical even in the absence of sophisticated analysis techniques. Throughout the first third of the century, ordinary carbon steel, without any special alloy strengthening or hardening, was universally used.The possibilities inherent in metal construction for high-rise building was demonstrated to the world by the Paris Exposition of 1889.for which Alexandre-Gustave Eiffel, a leading French bridge engineer, erected an openwork metal tower 300m (984 ft) high. Not only was the height-more than double that of the Great Pyramid-remarkable, but the speed of erection and low cost were even more so, a small crew completed the work in a few months.The first skyscrapers. Meantime, in the United States another important development was taking place. In 1884-85 Maj. William Le Baron Jenney, a Chicago engineer , had designed the Home Insurance Building, ten stories high, with a metal skeleton. Jenney’s beams were of Bessemer steel, though his columns were cast iron. Cast iron lintels supporting masonry over window openings were, in turn, supported on the cast iron columns. Soild masonry court and party walls provided lateral support against wind loading. Within a decade the same type of construction had been used in more than 30 office buildings in Chicago and New York. Steel played a larger and larger role in these , with riveted connections for beams and columns, sometimes strengthened for wind bracing by overlaying gusset plates at the junction of vertical and horizontal members. Light masonry curtain walls, supported at each floor level, replaced the old heavy masonry curtain walls, supported at each floor level , replaced the old heavy masonry.Though the new construction form was to remain centred almost entirely in America for several decade, its impact on the steel industry was worldwide. By the last years of the 19th century, the basic structural shapes-I beams up to 20 in. ( 0.508m) in depth and Z and T shapes of lesser proportions were readily available, to combine with plates of several widths and thicknesses to make efficient members of any required size and strength. In 1885 the heaviest structural shape produced throughhot-rolling weighed less than 100 pounds (45 kilograms) per foot; decade by decade this figure rose until in the 1960s it exceeded 700 pounds (320 kilograms) per foot.Coincident with the introduction of structural steel came the introduction of the Otis electric elevator in 1889. The demonstration of a safe passenger elevator, together with that of a safe and economical steel construction method, sent building heights soaring. In New York the 286-ft (87.2-m) Flatiron Building of 1902 was surpassed in 1904 by the 375-ft (115-m) Times Building ( renamed the Allied Chemical Building) , the 468-ft (143-m) City Investing Company Building in Wall Street, the 612-ft (187-m) Singer Building (1908), the 700-ft (214-m) Metropolitan Tower (1909) and, in 1913, the 780-ft (232-m) Woolworth Building.The rapid increase in height and the height-to-width ratio brought problems. To limit street congestion, building setback design was prescribed. On the technical side, the problem of lateral support was studied. A diagonal bracing system, such as that used in the Eiffel Tower, was not architecturally desirable in offices relying on sunlight for illumination. The answer was found in greater reliance on the bending resistance of certain individual beams and columns strategically designed into the skeletn frame, together with a high degree of rigidity sought at the junction of the beams and columns.With today’s modern interior lighting systems, however, diagonal bracing against wind loads has returned; one notable example is the John Hancock Center in Chicago, where the external X-braces form a dramatic part of the structure’s façade.World War I brought an interruption to the boom in what had come to be called skyscrapers (the origin of the word is uncertain), but in the 1920s New York saw a resumption of the height race, culminating in the Empire State Building in the 1931. The Empire State’s 102 stories (1,250ft. [381m]) were to keep it established as the hightest building in the world for the next 40 years. Its speed of the erection demonstrated how thoroughly the new construction technique had been mastered. A depot across the bay at Bayonne, N.J., supplied the girders by lighter and truck on a schedule operated with millitary precision; nine derricks powerde by electric hoists lifted the girders to position; an industrial-railway setup moved steel and other material on each floor. Initial connections were made by bolting , closely followed by riveting, followed by masonry and finishing. The entire job was completed in one year and 45 days.The worldwide depression of the 1930s and World War II provided another interruption to steel construction development, but at the same time the introduction of welding to replace riveting provided an important advance.Joining of steel parts by metal are welding had been successfully achieved by the end of the 19th century and was used in emergency ship repairs during World War I, but its application to construction was limited until after World War II. Another advance in the same area had been the introduction of high-strength bolts to replace rivets in field connections.Since the close of World War II, research in Europe, the U.S., and Japan has greatly extended knowledge of the behavior of different types of structural steel under varying stresses, including those exceeding the yield point, making possible more refined and systematic analysis. This in turn has led to the adoption of more liberal design codes in most countries, more imaginative design made possible by so-called plastic design ?The introduction of the computer by short-cutting tedious paperwork, made further advances and savings possible.中文翻译高层结构与钢结构近年来，尽管一般的建筑结构设计取得了很大的进步，但是取得显著成绩的还要属超高层建筑结构设计。

土木专业毕业设计外文翻译英文文献

144 Study on Construction Cost of Construction ProjectsHui LiAudit Department of Tianjin Polytechnic UniversityE-mail: lihui650122@AbstractChina is a country which has the largest investment amount in engineering construction in the world and which has the most construction projects. It is a significant subject for the extensive engineering managers to have effective engineering cost management in construction project management and to reasonably determine and control construction cost on the condition of ensuring construction quality and time limit.On the basis of the status quo of losing control in Chinese construction investment and of separation of technique and economy in engineering, and guided by basic theories of construction cost control, the author discusses control methods and application of construction cost, sets forth existing issues in construction cost control and influences of these issues on determination and control of construction cost, puts forward that construction cost control should reflect cost control of the entire construction process at the earlier stage of construction, and then introduces some procedures and methods of applying value project cost control at all stages of construction projects.Keywords: Construction cost, Cost control, Project1. Significance of the studyThe existing construction cost management system in China was formulated in 1950s, and improved in 1980s. Traditional construction cost managerial approach was one method brought in from the former Soviet Union based on unified quota of the country. It is characterized by the managerial approach of construction cost in the planned economy, which determines that it cannot adapt to requirements of the current market economy.Traditional construction cost managerial approach in China mainly includes two aspects, namely, determination approach of construction cost and control approach of construction cost. The traditional determination approach of construction cost mainly applied mechanically national or local unified quantity quota to determine the cost of a construction project. Although this approach has undergone reform of over 20 years, until now, influences of planned economy management mode have still been in existence in many regions. Control approach of our traditional construction cost is mainly to control settlement and alteration of construction cost, which is merely an approach to settle accounts after the event, and which cannot satisfy the purpose of saving resources and improving work. In recent years, requirements of developed countries on project investment have been to plan to control in advance and to control in the middle of an event, whose effects have proved to be effective. An actually scientific approach should be that construction cost control approach beforehand and after the event can eliminate or diminish labour in vain or poor efficiency and unnecessary resource degradation and methods applied in implementation of construction projects before or after the event.Considering the above situation, the academic circles put forward concept of cost management and control of the overall process as early as 1980s. They began to attach importance to prophase management of construction projects and take the initiative to conduct cost management. Afterwards, on July 1, 2003, implementation of <<Cost Estimate Norm for Bill of Quantity of Construction Works>> symbolized that cost estimate of China had entered a brand-new era that complied with development rules of market economy. From then on, concepts and approaches of Chinese cost management were really integrated with the international society.Losing control of construction project investment is a universal phenomenon in fixed investment field in China. A construction project consumes quite a lot of manpower, materials and machines, with large investment, long construction cycle, and strong synthesis, so it is related with economic interests of all construction parties and means a lot to national economy. Currently, in the field of Chinese project construction, there exists the status quo of separation of technique and economy. Most of engineers and technicians tend to regard construction cost as duty of financing andpreliminary budget personnel, and mistakenly believe that it has nothing to do with themselves. In the process of carrying out a project, they usually only focus on quality control and progress control, while they ignore control over investment in construction projects. If technicians ignore construction cost, and those who are in charge of construction cost have no knowledge in relevant technical construction connected with construction cost, then it is difficult for them to reasonably confirm and effectively control construction cost. Construction supervision investment control refers to managerial activities at the whole implementation state of the project, which attempts to guarantee realization of project investment targets with the premise of satisfying quality and progress. Investment targets are set at different stages with further progress of construction practice, and construction cost control runs through the entire process of project construction, but it should give prominence to the key points. Obviously, the key of construction cost control lies in investment decision-making and design stage before the construction, while after the investment decision is made, the key lies in the design. Life cycle of construction project includes construction cost and recurrent expenditure after the construction project is put into service, and discard and removal costs etc after usage period of the project. According to analysis of some western countries, usually design cost only amounts to less than 1% of life cycle of construction project. However, it is the cost of less than 1% that accounts for more than 75% of influences on construction cost. It is therefore obvious that, design quality is vital to benefits of the entire project construction.For a long time, construction cost control of the preliminary engineering of project construction has been ignored in China, while the primary energy of controlling construction cost has been focused upon auditing working drawing estimate, settling construction cost and settling itemized account during construction. Although this has its effect, after all, this had no difference from taking precautions after suffering a loss and getting half the result with twice the efforts. In order to effectively control construction cost, the emphasis of control should be firmly transferred to preliminary construction stage. At present, we should take all pains to grasp this significant stage so as to achieve maximum results with little effort.This article aims to analyze existing issues in cost control of the entire construction period through study on theoretical methods and practice of construction cost management. Especially, issues in cost control in the earlier period of construction deserve our research, so that we can explore corresponding reform measures to offer some references for construction project cost control.The situation of a construction project in which budgetary estimate exceeds estimation, budget exceeds budgetary estimate, and settlement exceeds budget, is a universal phenomenon in investment in fixed assets in China. Construction cost which is out of control adds to investment pressure, increases construction cost, reduces investment profit, affects investment decision-making, and, to a great extent, wastes the national finance, so it is likely to result in corruption or offence. Since the middle of 1950s, on the basis of summarizing practical experiences of fundamental construction battle line for several decades, we have conducted a series of reforms in construction field. Especially since May 1988, we have gradually implemented the system of construction supervision all over the country, which has had some positive effects upon reversing the phenomenon of losing control of a construction project in the implementation period. However, because that system is still in its starting stage, there hasn’t appeared a large batch of professional and socialized supervision teams. In addition, in projects in which construction supervision is carried out, there exist general phenomena, such as “emphasis on quality control at the construction stage and neglect of investment control”, and “emphasis on technical aspects of supervision and neglect of economic aspects of supervision”. In reality, rights of supervision tend to be confined to management of technical aspects, while management of economic aspects is firmly in control of proprietors. Meanwhile, lagging behind of existing construction cost management system is the primary cause for losing control of construction cost. Therefore, as a whole, the phenomenon of losing control over construction project cost is still quite serious, so it is necessary to conduct further study and make further analysis on major factors of current construction cost management and factors at all stages of a construction project that affect construction cost.2. Primary study contentAiming at the subject of “control of whole-process of construction project cost”, and based on lots of literature reviews about determination and control of construction project cost both at home and abroad, the author of this paper has collected extensively some relevant provincial and city reports and data after investigation. Afterwards, the author conducts the following work.1) To analyze formulation of construction project investment and to find out primary reasons for losing control over construction cost at all stages of a construction project.2) To study and analyze status quo and existing issues of current construction cost management, and study influences of these issues upon determination and control of a construction cost.3) To put forward effective approaches and methods as well application of value engineering of a construction project from its decision-making stage, design stage, construction stage to the final acceptance of construction stage.1454) To make clear significance, necessity and feasibility of cost control of a construction project so as to provide recommendations for improvement of construction cost management in China.2.1 Construction cost control theory and management mod eAccording to the new cost control theory, cost engineers are “professional persons who undertake cost estimate, cost control, marketing planning and scientific management”. Fields undertaken by cost engineers include such aspects as project management, project planning, progress management and profitability analysis etc of a project construction and its production process. Cost engineers offer service for control over life cycle expenditure, property facilities and production & manufacture of a construction project with their management technique with an overall cost.2.2 Current construction cost management model and theories in China2.2.1 Direct regulation and control of the governmentConsidering development process of quota, it can be discovered that quota has come into being, developed and become mature gradually with development of planned economy after foundation of PRC. Since China has carried out centralized management model of investment system for a long time, the government is not only a maker of macropolicy, but a participant of micro-project construction. Therefore, a unified quota with dense colour of planned economy is able to provide powerful methods and means for the government to carry out macro-investment regulation and control and micro-construction project management.2.2.2 Valuation basis for current construction costBasic materials for calculation of construction cost usually include construction cost quota, construction cost expense quota, cost index, basic unit price, quantities calculation rule and relevant economic rules and policies issued by competent departments of the government, etc. It includes index of estimate (budgetary estimate index), budgetary estimate quota, budgetary quota (comprehensive budgetary quota), expense quota (standard), labor quota, working-day norm, materials, budgetary price of facilities, direct price index of a project, material price index and cost index. And also included is valuation criterion of consumption quota and list of items in recent two years.2.2.3 Valuation model of current construction costValuation model is a basic aspect of construction cost management. Construction cost management is a governmental behavior, while valuation model is a means for a country to manage and control construction cost. There are two construction valuation models at present in China, namely, valuation model according to quota and one according to bill of quantities.2.2.3.1 Valuation model according to quotaValuation model according to quota is an effective model adopted during the transition period from planned economy to market economy. Determination of construction cost through valuation model according to quota prevents overrated valuation and standards and prices pressed down to some extent, because budgetary quota standardizes rate of consumption and a variety of documents stipulate manpower, materials, unit price of machines and all sorts of service fee norms, which reflects normativity, unitarity and rationality of construction cost. However, it has an inhibited effect upon market competition, and is not favorable for a construction enterprise to improve its technique, strengthen its management and enhance its labor efficiency and market competition.2.2.3.2 Valuation model according to bill of quantitiesValuation model according to bill of quantities is a construction cost determination model proposed recently. In this model, the government merely unifies project code, project name, unit of measurement and measurement rule of quantities. Each construction enterprise has its self-determination to quote a price according to its own situation in a tender offer, and price of building products is formed thereby in the process of bidding.2.3 Cost control in the process of implementationFor a long time, technique and economy has been separated in the field of project construction. Restrained by the planned economy, there lacks the economic concept in the minds of our engineers and technicians, because they regard reduction of construction cost as a duty of financial personnel which has nothing to do with themselves. However, the primary responsibility of financial and preliminary budget personnel is to act in accordance with financial system. Usually, they are not familiar with construction technique, and know little or even nothing about changes of various relations in project design, construction content and implementation of construction. Under such a circumstance, they have no choice but to mechanically work out or audit the expenditure from a financial perspective, which results in mutual separation of technique and economy. They just do what they do, which negatively reflects price of quantities of a project that has been completed, so it is difficult to control construction cost rationally and effectively.1462.4 Control of cost in the process of constructionImplementation stage of a construction project is a stage which requires the most assets in the whole process of a project construction, and is also a vital stage for pecuniary resources to transform into building entities. Cost control at the implementation stage refers to confine construction cost within a scheduled control scope through a scientific cost control theory and method on the condition of ensuring project quality and time limit. The process of generation of a building entity is inreversible, so if effective automatic control and precontrol cannot be conducted over construction cost, then economic loss might be caused that cannot be made up for.2.5 Analysis of major factors that affect construction cost at the stage of implementationImplementation stage of a project refers to the period from completion of construction documents design and examination and submission to the construction party to the final completion acceptance of the project and until it is put into use. According to the basic operation procedure of the implementation stage of a construction project, formation of a construction cost has to undergo such major aspects as bidding, contract signing and management, joint auditing of a shop drawing, investigation of a construction management plan, material management and completion settlement, etc. All these aspects affect construction cost settlement to different degrees. In that process, after evolving from budgetary price, price for successful bidding, refurbishing cost for a contract, the construction cost is finally determined in the form of settlement price for project completion. Factors affecting construction cost are various, but from the perspective of analysis of cost formation, there are primarily the following reasons.1) Influences of a project bidding. Bidding can determine price for successful bidding, while contract price is determined on the basis of price for successful bidding. If something goes wrong with bidding, then it might result in distortion of the price for bidding, and it is impossible to provide accurate and reliable foundation for cost control, and even result in losing control over the cost.2) Influences of contract signing and management. Determination of a contract price further makes precise target of cost control, and an initial draft of a contract term provides correct foundation and principles for cost control. After signing of a contract, contract items are regarded as foundation, which will have strict contract control over design changes at the construction stage, project measurement, payment of a construction debt, and construction compensation, etc, and which will ensure realization of a control target. Therefore, losing control over signing and management of a contract will necessarily result in losing control over construction cost.3) Influences of examination of construction management plan. Construction management plan is one of important foundations for determine a project bidding price and contract price. In the process of construction, adjustment of a contract price should also be determined according to construction management plan, because quality of construction management plan will directly affect quality and progress of a project. Therefore, losing control over examination of construction management plan will bring extremely unfavorable influences upon control over construction cost.4) Influences of material management. On one hand, material price is an important component of bidding price and contract price. On the other hand, material expense accounts for a large proportion in construction cost, because price of materials determine construction cost. Therefore, losing control over material management will necessarily result in losing control over construction cost.5) Influences of settlement, examination and verification of a project completion. Settlement, examination and verification is the final stage of a construction cost control at the implementation stage. A strict and meticulous settlement, examination and verification can ensure accuracy and authenticity of settlement cost of a project. According to previous analysis, we believe that all aspects of cost control can have effect upon formation of construction cost, among which bidding of a project, contract signing and management, examination of a construction management plan and management of materials all have decisive effects upon formation of construction cost, and are vital aspects in cost construction at the implementation stage of a project, so neglect of these four aspects is a direct cause for losing control over construction cost.In this paper, the author summarizes relevant issues in construction cost control at the decision-making stage of a construction project, at the design stage and construction stage, and puts forward principles or resolutions for handing such issues. Especially, as a method of combination of technique and economics, application of value engineering is elaborated at all stages, so that construction cost gets effective controlled. This paper cannot conclude all such issues existing, and also resolutions to resolve these issues cannot cover and contain everything, but with development of construction, new issues and new trains of thought will continue to emerge.ReferencesAminan Fayek. (1998). Competitive Bidding Strategy Model and Software System for Bid Preparation. Jounal of Construction Engineering and Management.Chen, Jianguo. (2001). Project Measurement and cost management. Shanghai: Tongji University Press.147Don R.Hansen & Maryanne M. Mowen. (2005). Cost Management: Accounting and Control.Dong, Shibo. (2003). Status Quo of Construction Cost Management Theory and Its Developmental Trend. Construction Cost Management, (5).Feng, Jingchun. (2000). Study on Counter Measures of Project Cost Management. Technical and Economic Development, (6).George J.Ritz. (1993). Total Construction Project Management.Gou, Zhiyuan. (2002). Thought on Integrated Control Approach of Construction Cost Management. Construction Cost Management, (6).Hao, Jianxin. (2002). American Construction Cost Management. Tianjin: Nankai University Press, 1, 51.Hu, Jianming. (2002). Discussion on Construction Cost Estimation Consultant Participating in Whole Course of Cost Management. Construction Cost Management, (5).Hu, Zhifeng. (2000). Overall Process Control on Construction Projects. Coal Enterprise Management, (7).Huang, Yonggen. (2004). Value Engineering and Its Application in Construction Cost Control. Construction Economics, (8).Ivor H Seeley. (1996). Building economics (fourth edition). Macmillan Press LTD.James A.Bent & Kenneth King Humphreys. (1996). Effective Project Management through Applied Cost and Schedule Control, Cost Engineering.Jan Emblemsavg. (2003). Life cycle Costing: sing Activity-based Costing and Monte Carlo Methods to Manage Future Costs and Risks. John wiley & sons, (5).Janice T. Dana. (1999). Standardized Quantity Recipe File for Quality and Cost Control.John E.Schaufelberger & Len Holm. (2001). Management of Construction Projects: A Constructor's Perspective.John Innes, Falconer Mitchell & Takeo Yoshikawa. (2000). Activity Costing for Engineers. Research Studies Press Ltd. John R.Canada, William G Sullivan, Dennis 3. Kulonda & John A.White. (2004). Capital Investment Analysis for Engineering and Management.Li, Tinggui. (2003). Study on Cost Management Model and Countermeasures of Construction project after China's entry into the WTO. Construction Cost Management.Liu, Guiwen & Shen, Qiping. (2001). A Study of Value Engineering Applications in China’s Construction Industry. Value Engineering, (3).Liu, Hongqing. (2003). About overall cost control. Shanxi Architecture, (29)6.Liu, Zhongying & Mao, Jian. Architecture Project Quantity List Quotation. Southeast University Press, 9.Luo, Dinglin. (1997). Determination and Control of Construction Project Cost at Home and Abroad. Beijing: Chemical Industry Press.Ma, Guanghong & Xu, Wei. (2003). Discussion on Application of Overall Cost Management Theory. Project Management, (4).Ma, Guanghong & Xu, Wei. (2003). Discussion on Application of Overall Cost Management Theory. Project Management, (4).Norton B R & McElligot C W. (1995). Value management in construction: a practical guide. Hampshire: Macmillan Press.Paul J. McVety. (1997). The Menu and the Cycle of Cost Control.Project Management Institute. (2004). A Guide to the Project Management Body of Knowledge.Qi, Anbang. (2000). Total Cost Management for Engineering Project. Tianjin: Nankai University Press.Qin, Aiguo. (1999). Study on Construction Cost Management. Economic Tribune, (22).Ren, Guoqiang & Yin, Yilin. (2003). The Feasibility Study on Life Cycle Cost Management in Terms of Paradigm Transformation. China Soft Science Magazine, (5).Ren, Hong. (2004). Cost Planning and Control of Construction Project. China Higher Education Press.Sidney M.Levy. (2002). Project Management in Construction.Stephen P Robbins & David A. Decenzo. (2002). Fundament of Management. Prentice Hall, Inc.148Takashi Ishikawa. (1996). Analogy by Abstraction: Case Retrieval and Adaptation for Inventive Design Expert Systems. Expert Systems with Application, (4)10.Tao, Xueming, Huang, Yunde & Xiong, Wei. (2004). Construction Cost Valuation and Management. China Architecture & Building Press, 2.Wang, Ailin. (2003). Value Engineering and Its Application in Constructional Engineering. Anhui Architecture, (5). Wang, Li & Xu, Zihua. (2004). Comparative Study on Construction Cost Models at Home and Abroad. Architecture Economics.Wang, Yulong. (1997). 2000 Cases on Issues of Construction Project Cost. Shanghai: Tongji University Press. Wang, Zhenqiang. (2002). British Construction Cost Management. Tianjin: Nankai University Press.Wang, Zhenqiang. (2002). Japanese Construction Cost Management. Tianjin: Nankai University Press, 4.Xiang, Ke & Luo, Feng. (2004). Cost Control of Design Stage. Sichuan Architecture, (2).Xu, Datu. (1997). Determination and Control of Construction Cost. Beijing: China Planning Press.Xu, Datu. (1997). Investment Control of Construction Project. Beijing: China Planning Press.Yin, Yilin. (2001). Determination and Control of Construction Cost. Beijing: China Planning Press.Zhang, Caijiang, Li, Kehua & Xu, Yongmei. Review of VE Theory and Practice in China and Some Deep Thinking about its Depression. Nankai Business Review, (1).Zhong, Guangen. (2004). Brief Discussion on Cost Control System in Projects of Commonwealth Nations.Zuo, Jin & Han, Hongyun. Actuality & Amelioration of Whole Life-cycle Value-chain in Architecture. Value Engineering, (6).149。

土木工程建筑工程毕业设计外文翻译

大连交通大学2011届本科生毕业设计（论文）外文翻译Seismic Collapse Safety of Reinforced Concrete Buildings:I. Assessment of Ductile Moment FramesCurt B. Haselton1, Abbie B. Liel2, Gregory G. Deierlein3, Brian S. Dean4, Jason H. Chou5Ground motions used for the nonlinear dynamic analyses are recordings from large magnitude earthquakes (magnitude 6.5 to 7.6) recorded at moderate fault rupturedistances (10 to 45 km). The 39 ground motion record pairs (each with two orthogonal horizontal components) and their selection criteria are documented in Haselton and Deierlein (2007). This ground motion set is an expanded version of the far-field ground motion set utilized in the FEMA P-695 (FEMA 2009).Ground motion records are selected and scaled without considering the distinctive spectral shape of rare (extreme) ground motions, due to difficulties in selecting and scaling a different set of records for a large set of buildings having a wide range of first- mode periods. To account for the important impact of spectral shape on collapse assessment, shown by Baker and Cornell (2006), the collapse predictions made using the general set of ground motions are modified using a method proposed by Haselton et al. (2009). The expected spectral shape of rare (large) California ground motions isaccounted for through a statistical parameter referred to as epsilon (ε), which is a measure of the difference between the spectral acceleration of a recorded ground motion and the median value predicted by ground motion prediction equation. A target value of ε=1.5 is used to approximately represent the expected spectral shape of severe ground motions that can lead to collapse of code-conforming buildings (Appendix B of FEMA P-695 2009; Haselton et al. 2010).Page 1 of 7大连交通大学2011届本科生毕业设计（论文）外文翻译STRUCTURAL ANALYSIS MODEL AND COLLAPSE ASSESSMENT METHODOLOGYA two-dimensional three-bay nonlinear analysis frame model is created for each archetype RC SMF using the OpenSees structural analysis platform (OpenSees 2009), as illustrated in Figure 1. Three bays are assumed to be the minimum number necessary to reflect the differences between interior and exterior columns and joints, and their impact on frame behavior. Strength and stiffness of the gravity system are not represented in the model, but the destabilizing P-Δ effectsare accounted for by applying gravity loads on a leaning column in the analysis model. Previous research by the authors has shown that neglecting the strength and stiffness of the gravity system in RC SMF systems is slightly conservative, underestimating the median collapse capacity by approximately 10% (Haselton et al. 2008a). It is also assumed that the damage to the slab-column connections of the gravity system will not result in a vertical collapse of the slab; test data for slab-column connections with modern detailing are still needed to verify this assumption. The foundation rotation stiffness is calculated from typical grade beam design and soil stiffness properties. Rayleigh damping corresponding to 5% of critical damping in the first and third modes is applied.Element modeling consists of lumped plasticity beam-column elements and finite joint shear panel springs. Lumped plasticity elements were used in lieu of fiber-type element models, since only the former are able to capture the strain softening associated with rebar buckling and spalling phenomena that are critical for simulating structural collapse in RC frame structures. The beam-columns are modeled using a nonlinear hinge model with degrading strength and stiffness, developed by Ibarra et al. (2005). As illustrated in Figure 2, the Ibarra et al. model captures the important modes of monotonicPage 2 of 7大连交通大学2011届本科生毕业设计（论文）外文翻译and cyclic deterioration that precipitate sidesway collapse. Key parameters of the modelinclude the plastic rotation capacity, θcap,pl, the post-capping rotation capacity, θpc, theratio of maximum to yield moment, Mc / My, and an energy-based degradation parameter,. Based on calibration to test data for RC columns and beams with ductile detailing andlow to moderate axial load, the typical mode parameter values are θcap,pl between 0.035 to0.085 radians, depending on the level of axial load in the beam-column, θpc equal to 0.10radians, Mc / My between 1.17 and 1.21, and between 85 and 130 (Haselton et al. 2007,2008b). The post-capping deformation capacity, θpc, of 0.10 is a conservative value used dueto lack of data; this value would likely be much larger if additional data were availablewith specimens tested to larger levels of deformation.The collapse capacities of the archetype building designs are evaluated using aperformance-based methodology, key features of which are briefly summarized as follows:1. Select ground motions for nonlinear dynamic analysis. In this study, 39 pairs offar-field ground motions are used. Issues related to record selection and scaling have been discussed previously.2. Utilize incremental dynamic analysis (IDA) to organize nonlinear dynamiccollapse analyses of the archetype models subjected to the recorded ground motions (Vamvatsikos and Cornell 2002). Using the IDA approach, each horizontal component of ground motion is individually applied to the two-dimensional frame model.In this study, ground motion records are amplitude scaled according to thespectral acceleration at the first mode period, Sa(T1). The ground motions are increasingly scaled until collapse occurs. In this paper, collapse is defined as the Page 3 of 7大连交通大学2011届本科生毕业设计（论文）外文翻译point of dynamic instability, where the lateral story drifts of the building increase without bounds (often referred to as sidesway collapse). This occurs when the IDA curve becomes flat. Vertical collapse mechanisms, which are not directly simulated in the structural model, are not considered in this assessment. The companion paper (Liel et al. 2010) provides explanation for how these additional collapse modes but could be accounted for.Figure 3a presents sample results from incremental dynamic analysis for a four-story space frame building (ID1008). For this structure, the median collapse capacity (in terms of Sa(0.94s)) is 1.59g for the set of 39 ground motion pairs.3. Construct a collapse fragility function based on the IDA results, which represents the probability of collapse as a function of ground motion intensity. To approximately account for three-dimensional ground motion effects (i.e. themaximum ground motion component), the lower collapse capacity (in terms of Sa(T1)) from each pair of motions is recorded as the building collapse capacity. From the resulting collapse data, the median collapse capacity and dispersion, due to record-to-record variability, are then computed.Figure 3b presents such collapse fragility curves for the four-story building usedpreviously in Figure 3a. The square markers show the empirical cumulative distribution function of the collapse data from Figure 3a (i.e. each point represents the collapse capacity for a single earthquake record), and the solid line shows the lognormal distribution fit to the empirical data. The fitted median collapse capacity (in terms of Sa(0.94s)) is 1.59g and the fitted logarithmic standardPage 4 of 7大连交通大学2011届本科生毕业设计（论文）外文翻译deviation, representing the so-called record-to-record (RTR) variability (LN,RTR), is 0.38.4. Increase the dispersion in the collapse fragility to account for structural modeling uncertainties.Figure 3b shows this adjusted collapse capacity distribution by the dashed line. Liel et al. (2009) and Haselton and Deierlein (2007) have shown how introducing this additional dispersion in the collapse fragility can approximately account for the effects of uncertainties in the structural modeling parameters, but this approximation is only suitable for collapse probabilities in the lower tail of the fragility curve (Liel et al. 2009). Based on uncertainties in the nonlinearcomponent models (e.g., the capping rotation and post-peak softening slope shown in Figure 2), the modeling uncertainty is calculated as σLN,modeling = 0.5 (Haselton and Deierlein 2007). When combined with the record-to-record uncertainty of LN,RTR = 0.38, the resulting total dispersion is LN,total = 0.63, shown by the dashed curve labeled RTR+Model.5. Adjust (increase) the median of the collapse fragility curve to account for the ground motion spectral shape effect.Figure 3b shows this adjusted collapse capacity distribution by the dotted line. For this example, the median collapse intensity is increased from 1.59g to 2.22g (by a factor of 1.4). As described by Haselton et al. (2010) and FEMA P-695 (FEMA 2009, Appendix B), this so-called ε adjustment is based on the large ductility of the RC SMF structures and associated period shift that occurs before collapse, combined with a target value of ε = 1.5 for rare ground motions in thePage 5 of 7大连交通大学2011届本科生毕业设计（论文）外文翻译high seismic regions of California. Buildings with lower deformation capacity, as well as sit es and hazard levels with lower expected values of ε, would have a smalleradjustment.6. Compute the collapse risk metrics of interest.For the example in Figure 3b, the collapse margin ratio is 2.6, the conditional collapse probability (P(C|Sa2/50)) is 7%, and the mean annual frequency ofcollapse (λcol) is 1.7x10-4 collapses/year.COLLAPSE RISK FOR RC SMF BUILDINGS DESIGNED ACCORDING TO ASCE 7-02Collapse analysis results for the 30 building archetypes are summarized in Table 1. Pertinent data includes the fundamental period of each archetype structural model, static overstrength from pushover analysis, collapse risk predictions, and maximum story and roof drifts at the onset of collapse. The resulting collapse risks are described by the following three measures, as listed in Table 1 and plotted in Figure 4: Collapse Margin: The collapse margin is the ratio between the median collapse capacity and the 2% in 50 year ground motion level. This metric is similar in concept to a simple factor of safety. Overall, the collapse margins for the 30 RC SMF buildings range from 1.7 to 3.4, with an average value of 2.3.Conditional Collapse Probability: The probability of collapse for the 2% in 50 year level of ground motion intensity, denoted P(C|Sa2/50), can be read directly from the fragility curve. This is a convenient metric to gauge the collapse safety relative to the extreme ground motion intensity that is used as the basis of design in building codes . Overall, the RC SMF buildings have an average P(C|Sa2/50) of 11%, with a range from 3% to 20%.Page 6 of 7大连交通大学2011届本科生毕业设计（论文）外文翻译Mean Annual Frequency of Collapse: The mean annual frequency of collapse (λcol) is obtained by integrating the collapse fragility with the site-specific hazard curve. Using the hazard curve from the Los Angeles site, the RC SMF buildings have an average λcol of 3.1x10-4 collapses/year, with a range from 0.7x10-4 to7.0x10-4 collapses/year. This range translates to a probability of collapse in 50 years of 0.4% to 3.4%.While there are no clear standards that define the maximum acceptable collapse risk for buildings, there is some consensus that calculated values for the RC SMF archetypes are in a reasonable range. For example, the FEMA P-695 (FEMA 2009) Methodology to determine seismic response factors for new building systems, is based on a maximum collapse risk of 10% to 20%, conditioned on the maximum considered earthquakeintensity. Additionally, the ASCE/SEI 7 building code has recently adopted new “risk consistent” seismic design maps, which have an implied collapse risk of 1% in 50 years (Luco et al. 2007), and which were developed based on an assumed collapse probability of 10%, conditioned on the maximum considered earthquake intensity. Finally, it is important to remember that the collapse risks reported herein were calculated from archetype designs that conform to current building code provisions. So, to the extent that the evolution of building codes reflects societal values, the calculated collapse risks have legitimacy implicit in the natural progression of building codes and standards.Page 7 of 7钢筋混凝土建筑的抗震安全设计大连交通大学2011届本科生毕业设计（论文）外文翻译I.延性框架的分析Curt B. Haselton1, Abbie B. Liel2, Gregory G. Deierlein3, Brian S. Dean4, Jason H. Chou5应用于非线性动态分析的地面运动是中等深度(10 到45 千米)断层错动引起的震级为6.5至7.6的大地震。

土木工程建筑工程毕业设计外文翻译

土木工程毕业设计外文翻译最终中英文

7 Rigid-Frame StructuresA rigid-frame high-rise structure typically comprises parallel or orthogonally arranged bents consisting of columns and girders with moment resistant joints. Resistance to horizontal loading is provided by the bending resistance of the columns, girders, and joints. The continuity of the frame also contributes to resisting gravity loading, by reducing the moments in the girders.The advantages of a rigid frame are the simplicity and convenience of its rectangular form.Its unobstructed arrangement, clear of bracing members and structural walls, allows freedom internally for the layout and externally for the fenestration. Rigid frames are considered economical for buildings of up to' about25 stories, above which their drift resistance is costly to control. If, however,a rigid frame is combined with shear walls or cores, the resulting structure is very much stiffer so that its height potential may extend up to 50 stories or more. A flat plate structure is very similar to a rigid frame, but with slabs replacing the girders As with a rigid frame, horizontal and vertical loadings are resisted in a flat plate structure by the flexural continuity between the vertical and horizontal components.As highly redundant structures, rigid frames are designed initially on the basis of approximate analyses, after which more rigorous analyses and checks can be made. The procedure may typically include the following stages:1. Estimation of gravity load forces in girders and columns by approximate method.2. Preliminary estimate of member sizes based on gravity load forces witharbitrary increase in sizes to allow for horizontal loading.3. Approximate allocation of horizontal loading to bents and preliminary analysisof member forces in bents.4. Check on drift and adjustment of member sizes if necessary.5. Check on strength of members for worst combination of gravity and horizontalloading, and adjustment of member sizes if necessary.6. Computer analysis of total structure for more accurate check on memberstrengths and drift, with further adjustment of sizes where required. This stage may include the second-order P-Delta effects of gravity loading on the member forces and drift..7. Detailed design of members and connections.This chapter considers methods of analysis for the deflections and forces for both gravity and horizontal loading. The methods are included in roughly the order of the design procedure, with approximate methods initially and computer techniques later. Stability analyses of rigid frames are discussed in Chapter 16.7.1 RIGID FRAME BEHAVIORThe horizontal stiffness of a rigid frame is governed mainly by the bending resistance of the girders, the columns, and their connections, and, in a tall frame, by the axial rigidity of the columns. The accumulated horizontal shear above any story of a rigid frame is resisted by shear in the columns of that story (Fig. 7.1). The shear causes the story-height columns to bend in double curvature with points of contraflexure at approximately mid-story-height levels. The moments applied to a joint from the columns above and below are resisted by the attached girders, which also bend in double curvature, with points of contraflexure at approximately mid-span. These deformations of the columns and girders allow racking of the frame and horizontal deflection in each story. The overall deflected shape of a rigid frame structure due to racking has a shear configuration with concavity upwind, a maximum inclination near the base, and a minimum inclination at the top, as shown in Fig.7.1.The overall moment of the external horizontal load is resisted in each story level by the couple resulting from the axial tensile and compressive forces in the columns on opposite sides of the structure (Fig. 7.2). The extension and shortening of the columns cause overall bending and associated horizontal displacements of the structure. Because of the cumulative rotation up the height, the story drift dueto overall bending increases with height, while that due to racking tends to decrease. Consequently the contribution to story drift from overall bending may, in. the uppermost stories, exceed that from racking. The contribution of overall bending to the total drift, however, will usually not exceed 10% of that of racking, except in very tall, slender,, rigid frames. Therefore the overall deflected shape of a high-rise rigid frame usually has a shear configuration.The response of a rigid frame to gravity loading differs from a simply connected frame in the continuous behavior of the girders. Negative moments are induced adjacent to the columns, and positive moments of usually lesser magnitude occur in the mid-span regions. The continuity also causes the maximum girder moments to be sensitive to the pattern of live loading. This must be considered when estimating the worst moment conditions. For example, the gravity load maximum hogging moment adjacent to an edge column occurs when live load acts only on the edge span andalternate other spans, as for A in Fig. 7.3a. The maximum hogging moments adjacent to an interior column are caused, however, when live load acts only on the spans adjacent to the column, as for B in Fig. 7.3b. The maximum mid-span sagging moment occurs when live load acts on the span under consideration, and alternate other spans, as for spans AB and CD in Fig. 7.3a.The dependence of a rigid frame on the moment capacity of the columns for resisting horizontal loading usually causes the columns of a rigid frame to be larger than those of the corresponding fully braced simply connected frame. On the other hand, while girders in braced frames are designed for their mid-span sagging moment, girders in rigid frames are designed for the end-of-span resultant hogging moments, which may be of lesser value. Consequently, girders in a rigid frame may be smaller than in the corresponding braced frame. Such reductions in size allow economy through the lower cost of the girders and possible reductions in story heights. These benefits may be offset, however, by the higher cost of the more complex rigid connections.7.2 APPROXIMATE DETERMINATION OF MEMBER FORCES CAUSED BY GRAVITY LOADSIMGA rigid frame is a highly redundant structure; consequently, an accurate analysis can be made only after the member sizes are assigned. Initially, therefore, member sizes are decided on the basis of approximate forces estimated either by conservativeformulas or by simplified methods of analysis that are independent of member properties. Two approaches for estimating girder forces due to gravity loading are given here.7.2.1 Girder Forces—Code Recommended ValuesIn rigid frames with two or more spans in which the longer of any two adjacent spans does not exceed the shorter by more than 20 %, and where the uniformly distributed design live load does not exceed three times the dead load, the girder moment and shears may be estimated from Table 7.1. This summarizes the recommendations given in the Uniform Building Code [7.1]. In other cases a conventional moment distribution or two-cycle moment distribution analysis should be made for a line of girders at a floor level.7.2.2 Two-Cycle Moment Distribution [7.2].This is a concise form of moment distribution for estimating girder moments in a continuous multibay span. It is more accurate than the formulas in Table 7.1, especially for cases of unequal spans and unequal loading in different spans.The following is assumed for the analysis:1. A counterclockwise restraining moment on the end of a girder is positive anda clockwise moment is negative.2. The ends of the columns at the floors above and below the considered girder are fixed.3. In the absence of known member sizes, distribution factors at each joint aretaken equal to 1 /n, where n is the number of members framing into the joint in the plane of the frame.Two-Cycle Moment Distribution—Worked Example. The method is demonstrated by a worked example. In Fig, 7.4, a four-span girder AE from a rigid-frame bent is shown with its loading. The fixed-end moments in each span are calculated for dead loading and total loading using the formulas given in Fig, 7.5. The moments are summarized in Table 7.2.The purpose of the moment distribution is to estimate for each support the maximum girder moments that can occur as a result of dead loading and pattern live loading.A different load combination must be considered for the maximum moment at each support, and a distribution made for each combination.The five distributions are presented separately in Table 7.3, and in a combined form in Table 7.4. Distributions a in Table 7.3 are for the exterior supports A andE. For the maximum hogging moment at A, total loading is applied to span AB with dead loading only on BC. The fixed-end moments are written in rows 1 and 2. In this distribution only .the resulting moment at A is of interest. For the first cycle, joint B is balanced with a correcting moment of - (-867 + 315)/4 = - U/4 assigned to M BA where U is the unbalanced moment. This is not recorded, but half of it, ( - U/4)/2, is carried over to M AB. This is recorded in row 3 and then added to the fixed-end moment and the result recorded in row 4.The second cycle involves the release and balance of joint A. The unbalancedmoment of 936 is balanced by adding -U/3 = -936/3 = -312 to M BA (row 5), implicitly adding the same moment to the two column ends at A. This completes the second cycle of the distribution. The resulting maximum moment at A is then given by the addition of rows 4 and 5, 936 - 312 = 624. The distribution for the maximum moment at E follows a similar procedure.Distribution b in Table 7.3 is for the maximum moment at B. The most severe loading pattern for this is with total loading on spans AB and BC and dead load only on CD. The operations are similar to those in Distribution a, except that the T first cycle involves balancing the two adjacent joints A and C while recording only their carryover moments to B. In the second cycle, B is balanced by adding - (-1012 + 782)/4 = 58 to each side of B. The addition of rows 4 and 5 then gives the maximum hogging moments at B. Distributions c and d, for the moments at joints C and D, follow patterns similar to Distribution b.The complete set of operations can be combined as in Table 7.4 by initially recording at each joint the fixed-end moments for both dead and total loading. Then the joint, or joints, adjacent to the one under consideration are balanced for the appropriate combination of loading, and carryover moments assigned .to the considered joint and recorded. The joint is then balanced to complete the distribution for that support.Maximum Mid-Span Moments. The most severe loading condition for a maximum mid-span sagging moment is when the considered span and alternate other spans and total loading. A concise method of obtaining these values may be included in the combined two-cycle distribution, as shown in Table 7.5. Adopting the convention that sagging moments at mid-span are positive, a mid-span total; loading moment is calculated for the fixed-end condition of each span and entered in the mid-span column of row 2. These mid-span moments must now be corrected to allow for rotation of the joints. This is achieved by multiplying the carryover moment, row 3, at the left-hand end of the span by (1 + 0.5 D.F. )/2, and the carryover moment at the right-hand end by -(1 + 0.5 D.F.)/2, where D.F. is the appropriate distribution factor, and recording the results in the middle column. For example, the carryover to the mid-span of AB from A = [(1 + 0.5/3)/2] x 69 = 40 and from B = -[(1+ 0.5/4)/2] x (-145) = 82. These correction moments are then added to the fixed-end mid-span moment to give the maximum mid-span sagging moment, that is, 733 + 40 + 82 = 855.7.2.3 Column ForcesThe gravity load axial force in a column is estimated from the accumulated tributary dead and live floor loading above that level, with reductions in live loading as permitted by the local Code of Practice. The gravity load maximum column moment is estimated by taking the maximum difference of the end moments in the connected girders and allocating it equally between the column ends just above and below the joint. To this should be added any unbalanced moment due to eccentricity of the girderconnections from the centroid of the column, also allocated equally between the column ends above and below the joint.第七章框架结构高层框架结构一般由平行或正交布置的梁柱结构组成，梁柱结构是由带有能承担弯矩作用节点的梁、柱组成。

土木工程外文翻译

附件2：外文原文（电子或复印件）Cyclic behavior of steel moment frame connections under varying axial load and lateral displacements Abstract: This paper discusses the cyclic behavior of four steel moment connections tested under variable axial load and lateral displacements. The beam specim- ens consisted of a reducedbeam section, wing plates and longitudinal stiffeners. The test specimens were subjected to varying axial forces and lateral displace- ments to simulate the effects on beams in a Coupled-Girder Moment-Resisting Framing system under lateral loading. The test results showed that the specim- ens responded in a ductile manner since the plastic rotations exceeded 0.03 rad without significant drop in the lateral capacity. The presence of the longitudin- al stiffener assisted in transferring the axial forces and delayed the formation of web local buckling.1. IntroductionAimed at evaluating the structural performance of reduced-beam section (RBS) connections under alternated axial loading and lateral displacement, four full-scale specimens were tested. These tests were intended to assess the performance of the moment connection design for the Moscone Center Exp- ansion under the Design Basis Earthquake (DBE) and the Maximum Considered Earthquake (MCE). Previous research conducted on RBS moment connections [1,2] showed that connections with RBS profiles can achieve rotations in excess of 0.03 rad. However, doubts have been cast onthe quality of the seismic performance of these connections under combined axial and lateral loading.The Moscone Center Expansion is a three-story, 71,814 m2 (773,000 ft2) structure with steel moment frames as its primary lateral force-resisting system. A three dimensional perspective illustration is shown in Fig. 1. The overall height of the building, at the highest point of the exhibition roof, is approxima- tely 35.36 m (116ft) above ground level. The ceiling height at the exhibition hall is 8.23 m (27 ft) , and the typical floor-to-floor height in the building is 11.43 m (37.5 ft). The building was designed as type I according to the requi- rements of the 1997 Uniform Building Code.The framing system consists of four moment frames in the East–West direct- ion, one on either side of the stair towers, and four frames in the North–South direction, one on either side of the stair and elevator cores in the east end and two at the west end of the structure [4]. Because of the story height, the con- cept of the Coupled-Girder Moment-Resisting Framing System (CGMRFS) was utilized.By coupling the girders, the lateral load-resisting behavior of the moment framing system changes to one where structural overturning moments are resisted partially by an axial compression–tension couple across the girder system, rather than only by the individual flexural action of the girders. As a result, a stiffer lateral load resisting system is achieved. The vertical element that connects the girders is referred to as a coupling link.Coupling links are analogous to and serve the same structural role as link beams in eccentrically braced frames. Coupling links are generally quite short, having a large shear- to-moment ratio.Under earthquake-type loading, the CGMRFS subjects its girders to wariab- ble axial forces in addition to their end moments. The axial forces in theFig. 1. Moscone Center Expansion Project in San Francisco, CAgirders result from the accumulated shear in the link.Fig 2. Analytical model of CGMRFNonlinear static pushover analysis was conducted on a typical one-bay model of the CGMRF. Fig. 2 shows the dimensions and the various sections of the model. The link flange plates were 28.5 mm ⋅ 254 mm (1 1/8 in ⋅ 10 in) and the web plate was 9.5 mm ⋅ 476 mm (3 /8 in ⋅ 18 3/4 in). The SAP 2000 computer program was utilized in the pushover analysis [5]. The frame was characterized as fully restrained(FR). FR moment frames are those frames for 1170which no more than 5% of the lateral deflections arise from connection deformation [6]. The 5% value refers only to deflection due to beam–column deformation and not to frame deflections that result from column panel zone deformation [6, 9].The analysis was performed using an expected value of the yield stress and ultimate strength. These values were equal to 372 MPa (54 ksi) and 518 MPa (75 ksi), respectively. The plastic hinges’ load–deformation behavior was approximated by the generalized curve suggested by NEHRP Guidelinesfor the Seismic Rehabilitation of Buildings [6] as shown in.Fig. 3. △y was calcu- lated based on Eqs. (5.1) and (5.2) from [6], as follows:P–M hinge load–deformation model points C, D and E are based on Table 5.4 from [6] for△y was taken as 0.01 rad per Note 3 in [6], Table 5.8. Shear hinge load- load–deformation model points C, D and E are based on Table 5.8 [6], Link Beam, Item a. A strain hardening slope between points B and C of 3% of the elastic slope was assumed for both models.The following relationship was used to account for moment–axial load interaction [6]:where MCE is the expected moment strength, ZRBS is the RBS plastic section modulus (in3), is the expected yield strength of the material (ksi), P is the axial force in the girder (kips) and is the expected axial yield force of the RBS, equal to (kips). The ultimate flexural capacities of the beam and the link of the model are shown in Table 1.Fig. 4 shows qualitatively the distribution of the bending moment, shear force, and axial force in the CGMRF under lateral load. The shear and axial force in the beams are less significant to the response of the beams as compared with the bending moment, although they must be considered in design. The qualita- tive distribution of internal forces illustrated in Fig. 5 is fundamentally the same for both elastic and inelastic ranges of behavior. Thespecific values of the internal forces will change as elements of the frame yield and internal for- ces are redistributed. The basic patterns illustrated in Fig. 5, however, remain the same.Inelastic static pushover analysis was carried out by applying monotonically increasing lateral displacements, at the top of both columns, as shown in Fig.6. After the four RBS have yielded simultaneously, a uniform yielding in the web and at the ends of the flanges of the vertical link will form. This is the yield mechanism for the frame , with plastic hinges also forming at the base of the columns if they are fixed. The base shear versus drift angle of the model is shown in Fig. 7 . The sequence of inelastic activity in the frame is shown on the figure. An elastic component, a long transition (consequence of the beam plastic hinges being formed simultaneously) and a narrow yield plateau characterize the pushover curve.The plastic rotation capacity, qp, is defined as the total plastic rotation beyond which the connection strength starts to degrade below 80% [7]. This definition is different from that outlined in Section 9 (Appendix S) of the AISC Seismic Provisions [8, 10]. Using Eq. (2) derived by Uang and Fan [7], an estimate of the RBS plastic rotation capacity was found to be 0.037 rad:Fyf was substituted for Ry•Fy [8], where Ry is used to account for the differ- ence between the nominal and the expected yield strengths (Grade 50 steel, Fy=345 MPa and Ry =1.1 are used).3. Experimental programThe experimental set-up for studying the behavior of a connection was based on Fig. 6(a). Using the plastic displacement dp, plastic rotation gp, and plastic story drift angle qp shown in the figure, from geometry, it follows that:And: in which d and g include the elastic components. Approximations as above are used for large inelastic beam deformations. The diagram in Fig. 6(a) suggest that a sub assemblage with displacements controlled in the manner shown in Fig. 6(b) can represent the inelastic behavior of a typical beam in a CGMRF.The test set-up shown in Fig. 8 was constructed to develop the mechanism shown in Fig. 6(a) and (b). The axial actuators were attached to three 2438 mm ×1219 mm ×1219 mm (8 ft ×4 ft ×4 ft) RC blocks. These blocks were tensioned to the laboratory floor by means of twenty-four 32 mm diameter dywidag rods. This arrangement permitted replacement of the specimen after each test.Therefore, the force applied by the axial actuator, P, can be resolved into two or thogonal components, Paxial and Plateral. Since the inclination angle of the axial actuator does not exceed 3.0 , therefore Paxial is approximately equal to P [4]. However, the lateral component, Plateral, causes an additional moment at the beam-to column joint. If the axial actuators compress the specimen, then the lateral components will be adding to the lateral actuator forces, while if the axial actuators pull the specimen, the Plateral will be anopposing force to the lateral actuators. When the axial actuators undergo axial actuators undergo a lateral displacement _, they cause an additional moment at the beam-to-column joint (P-△ effect). Therefore, the moment at the beam-to column joint is equal to:where H is the lateral forces, L is the arm, P is the axial force and _ is the lateral displacement.Four full-scale experiments of beam column connections were conducted. The member sizes and the results of tensile coupon tests are listed in Table 2 All of the columns and beams were of A572 Grade 50 steel (Fy 344.5 MPa). The actual measured beam flange yield stress value was equal to 372 MPa (54 ksi), while the ultimate strength ranged from 502 MPa (72.8 ksi) to 543 MPa (78.7 ksi).Table 3 shows the values of the plastic moment for each specimen (based on measured tensile coupon data) at the full cross-section and at the reduced section at mid-length of the RBS cutout.The specimens were designated as specimen 1 through specimen 4. Test specimens details are shown in Fig. 9 through Fig. 12. The following features were utilized in the design of the beam–column connection:The use of RBS in beam flanges. A circular cutout was provided, as illustr- ated in Figs. 11 and 12. For all specimens, 30% of the beam flange width was removed. The cuts were made carefully, and then ground smooth in a direct- tion parallel to the beam flange to minimize notches.Use of a fully welded web connection. The connection between the beam web and the column flange was made with a complete joint penetration groove weld (CJP). All CJP welds were performed according to AWS D1.1 Structural Welding CodeUse of two side plates welded with CJP to exterior sides of top and bottom beam flan- ges, from the face of the column flange to the beginning of the RBS, as shown in Figs. 11 and 12. The end of the side plate was smoothed to meet the beginning of the RBS. The side plates were welded with CJP with the column flanges. The side plate was added to increase the flexural capacity at the joint location, while the smooth transition was to reduce the stress raisers, which may initiate fractureTwo longitudinal stiffeners, 95 mm ×35 mm (3 3/4 in ×1 3/8 in), were welded with 12.7 mm (1/2 in) fillet weld at the middle height of the web as shown in Figs. 9 and 10. The stiffeners were welded with CJP to column flanges.Removal of weld tabs at both the top and bottom beam flange groove welds. The weld tabs were removed to eliminate any potential notches introduced by the tabs or by weld discontinuities in the groove weld run out regions.Use of continuity plates with a thickness approximately equal to the beam flange thickness. One-inch thick continuity plates were used for all specimens.While the RBS is the most distinguishing feature of these test specimens, thelongitudinal stiffener played an important role in delaying the formation of web local buckling and developing reliable connection performance4. Loading historySpecimens were tested by applying cycles of alternated load with tip displacement increments of _y as shown in Table 4. The tip displacement of the beam was imposed by servo-controlled actuators 3 and 4. When the axial force was to be applied, actuators 1 and 2 were activated such that its force simulates the shear force in the link to be transferred to the beam. The variable axial force was increased up to 2800 kN (630 kip) at 0.5_y. After that, this lo- ad was maintained constant through the maximum lateral displacement.maximum lateral displacement. As the specimen was pushed back the axial force remained constant until 0.5 y and then started to decrease to zero as the specimen passed through the neutral position [4]. According to the upper bound for beam axial force as discussed in Section 2 of this paper, it was concluded that P =2800 kN (630 kip) is appropriate to investigate this case in RBS loading. The tests were continued until failure of the specimen, or until limitations of the test set-up were reached.5. Test resultsThe hysteretic response of each specimen is shown in Fig. 13 and Fig. 16. These plots show beam moment versus plastic rotation. The beam moment is measured at the middle of the RBS, and was computed by taking an equiva-lent beam-tip force multiplied by the distance between the centerline of the lateral actuator to the middle of the RBS (1792 mm for specimens 1 and 2, 3972 mm for specimens 3 and 4). The equivalent lateral force accounts for the additional moment due to P–△ effect. The rotation angle was defined as the lateral displacement of the actuator divided by the length between the centerline of the lateral actuator to the mid length of the RBS. The plastic rotation was computed as follows [4]:where V is the shear force, Ke is the ratio of V/q in the elastic range. Measurements and observations made during the tests indicated that all of the plastic rotation in specimen 1 to specimen 4 was developed within the beam. The connection panel zone and the column remained elastic as intended by design.5.1. Specimens 1 and 2The responses of specimens 1 and 2 are shown in Fig. 13. Initial yielding occurred during cycles 7 and 8 at 1_y with yielding observed in the bottom flange. For all test specimens, initial yielding was observed at this location and attributed to the moment at the base of the specimen [4]. Progressing through the loading history, yielding started to propagate along the RBS bottom flange. During cycle 3.5_y initiation of web buckling was noted adjacent to the yielded bottom flange. Yielding started to propagate along the top flange of the RBS and some minor yielding along the middle stiffener.During the cycle of 5_y with the increased axial compression load to 3115 KN (700 kips) a severe web buckle developed along with flange local buckling. The flange and the web local buckling became more pronounced with each successive loading cycle. It should be noted here that the bottom flange and web local buckling was not accompanied by a significant deterioration in the hysteresis loops.A crack developed in specimen 1 bottom flange at the end of the RBS where it meets the side plate during the cycle 5.75_y. Upon progressing through the loading history, 7_y, the crack spread rapidly across the entire width of the bottom flange. Once the bottom flange was completely fractured, the web began to fracture. This fracture appeared to initiate at the end of the RBS，then propagated through the web net section of the shear tab, through the middle stiffener and the through the web net section on the other side of the stiffener. The maximum bending moment achieved on specimen 1 during theDuring the cycle 6.5 y, specimen 2 also showed a crack in the bottom flange at the end of the RBS where it meets the wing plate. Upon progressing thou- gh the loading history, 15 y, the crack spread slowly across the bottom flan- ge. Specimen 2 test was stopped at this point because the limitation of the test set-up was reached.The maximum force applied to specimens 1 and 2 was 890 kN (200 kip). The kink that is seen in the positive quadrant is due to the application of the varying axial tension force. The load-carrying capacity in this zone did notdeteriorate as evidenced with the positive slope of the force–displacement curve. However, the load-carrying capacity deteriorated slightly in the neg- ative zone due to the web and the flange local buckling.Photographs of specimen 1 during the test are shown in Figs. 14 and 15. Severe local buckling occurred in the bottom flange and portion of the web next to the bottom flange as shown in Fig. 14. The length of this buckle extended over the entire length of the RBS. Plastic hinges developed in the RBS with extensive yielding occurring in the beam flanges as well as the web. Fig. 15 shows the crack that initiated along the transition of the RBS to the side wing plate. Ultimate fracture of specimen 1 was caused by a fracture in the bottom flange. This fracture resulted in almost total loss of the beam- carrying capacity. Specimen 1 developed 0.05 rad of plastic rotation and showed no sign of distress at the face of the column as shown in Fig. 15.5.2. Specimens 3 and 4The response of specimens 3 and 4 is shown in Fig. 16. Initial yielding occured during cycles 7 and 8 at 1_y with significant yielding observed in the bottom flange. Progressing through the loading history, yielding started to propagate along the bottom flange on the RBS. During cycle 1.5_y initiation of web buckling was noted adjacent to the yielded bottom flange. Yielding started to propagate along the top flange of the RBS and some minor yielding along the middle stiffener. During the cycle of 3.5_y a severe web buckle developed along with flange local buckling. The flange and the web localbuckling bec- ame more pronounced with each successive loading cycle. During the cycle 4.5 y, the axial load was increased to 3115 KN (700 kips) causing yielding to propagate to middle transverse stiffener. Progressing through the loading history, the flange and the web local buckling became more severe. For both specimens, testing was stopped at this point due to limitations in the test set-up. No failures occurred in specimens 3 and 4. However, upon removing specimen 3 to outside the laboratory a hairline crack was observed at the CJP weld of the bottom flange to the column. The maximum forces applied to specimens 3 and 4 were 890 kN (200 kip) and 912 kN (205 kip). The load-carrying capacity deteriorated by 20% at the end of the tests for negative cycles due to the web and the flange local buckling. This gradual reduction started after about 0.015 to 0.02 rad of plastic rotation. The load-carrying capacity during positive cycles (axial tension applied in the girder) did not deteriorate as evidenced with the slope of the force–displacement envelope for specimen 3 shown in Fig. 17.A photograph of specimen 3 before testing is shown in Fig. 18. Fig. 19 is a Fig. 16. Hysteretic behavior of specimens 3 and 4 in terms of moment at middle RBS versus beam plastic rotation.photograph of specimen 4 taken after the application of 0.014 rad displacem- ent cycles, showing yielding and local buckling at the hinge region. The beam web yielded over its full depth. The most intense yielding was observed in the web bottom portion, between the bottom flange and the middlestiffener. The web top portion also showed yielding, although less severe than within the bottom portion. Yielding was observed in the longitudinal stiffener. No yiel- ding was observed in the web of the column in the joint panel zone. The un- reduced portion of the beam flanges near the face of the column did not show yielding either. The maximum displacement applied was 174 mm, and the maximum moment at the middle of the RBS was 1.51 times the plastic mom ent capacity of the beam. The plastic hinge rotation reached was about 0.032 rad (the hinge is located at a distance 0.54d from the column surface,where d is the depth of the beam).5.2.1. Strain distribution around connectionThe strain distribution across the flanges–outer surface of specimen 3 is shown in Figs. 20 and 21. The readings and the distributions of the strains in specimens 1, 2 and 4 (not presented) showed a similar trend. Also the seque- nce of yielding in these specimens is similar to specimen 3.The strain at 51 mm from the column in the top flange–outer surface remained below 0.2% during negative cycles. The top flange, at the same location, yielded in compression only.The longitudinal strains along the centerline of the bottom–flange outer face are shown in Figs. 22 and 23 for positive and negative cycles, respectively. From Fig.23, it is found that the strain on the RBS becomes several times larg- er than that near the column after cycles at –1.5_y; this is responsible for theflange local buckling. Bottom flange local buckling occurred when the average strain in the plate reached the strain-hardening value (esh _ 0.018) and the reduced-beam portion of the plate was fully yielded under longitudinal stresses and permitted the development of a full buckled wave.5.2.2. Cumulative energy dissipatedThe cumulative energy dissipated by the specimens is shown in Fig. 24. The cumulative energy dissipated was calculated as the sum of the areas enclosed the lateral load–lateral displacement hysteresis loops. Energy dissipation sta- rted to increase after cycle 12 at 2.5 y (Fig. 19). At large drift levels, energy dissipation augments significantly with small changes in drift. Specimen 2 dissipated more energy than specimen 1, which fractured at RBS transition. However, for both specimens the trend is similar up to cycles at q =0.04 radIn general, the dissipated energy during negative cycles was 1.55 times bigger than that for positive cycles in specimens 1 and 2. For specimens 3 and 4 the dissipated energy during negative cycles was 120%, on the average, that of the positive cycles.The combined phenomena of yielding, strain hardening, in-plane and out- of-plane deformations, and local distortion all occurred soon after the bottom flange RBS yielded.6. ConclusionsBased on the observations made during the tests, and on the analysis of theinstrumentation, the following conclusions were developed:1. The plastic rotation exceeded the 3% radians in all test specimens.2. Plastification of RBS developed in a stable manner.3. The overstrength ratios for the flexural strength of the test specimens were equal to 1.56 for specimen 1 and 1.51 for specimen4. The flexural strength capacity was based on the nominal yield strength and on the FEMA-273 beam–column equation.4. The plastic local buckling of the bottom flange and the web was not accompanied by a significant deterioration in the load-carrying capacity.5. Although flange local buckling did not cause an immediate degradation of strength, it did induce web local buckling.6. The longitudinal stiffener added in the middle of the beam web assisted in transferring the axial forces and in delaying the formation of web local buckling. How ever, this has caused a much higher overstrength ratio, which had a significant impact on the capacity design of the welded joints, panel zone and the column.7. A gradual strength reduction occurred after 0.015 to 0.02 rad of plastic rotation during negative cycles. No strength degradation was observed during positive cycles.8. Compression axial load under 0.0325Py does not affect substantially the connection deformation capacity.9. CGMRFS with properly designed and detailed RBS connections is areliable system to resist earthquakes.出自《工程索引》，The Engineering Index，简称EI。

(完整版)土木工程毕业设计外文文献翻译

外文文献翻译Reinforced ConcreteConcrete and reinforced concrete are used as building materials in every country. In many, including the United States and Canada, reinforced concrete is a dominant structural material in engineered construction. The universal nature of reinforced concrete construction stems from the wide availability of reinforcing bars and the constituents of concrete, gravel, sand, and cement, the relatively simple skills required in concrete construction, and the economy of reinforced concrete compared to other forms of construction. Concrete and reinforced concrete are used in bridges, buildings of all sorts underground structures, water tanks, television towers, offshore oil exploration and production structures, dams, and even in ships.Reinforced concrete structures may be cast-in-place concrete, constructed in their final location, or they may be precast concrete produced in a factory and erected at the construction site. Concrete structures may be severe and functional in design, or the shape and layout and be whimsical and artistic. Few other building materials off the architect and engineer such versatility and scope.Concrete is strong in compression but weak in tension. As a result, cracks develop whenever loads, or restrained shrinkage of temperature changes, give rise to tensile stresses in excess of the tensile strength of the concrete. In a plain concrete beam, the moments about the neutral axis due to applied loads are resisted by an internal tension-compression couple involving tension in the concrete. Such a beam fails very suddenly and completely when the first crack forms. In a reinforced concrete beam, steel bars are embedded in the concrete in such a way that the tension forces needed for moment equilibrium after the concrete cracks can be developed in the bars.The construction of a reinforced concrete member involves building a from of mold in the shape of the member being built. The form must be strong enough to support both the weight and hydrostatic pressure of the wet concrete, and any forces applied to it by workers, concrete buggies, wind, and so on. The reinforcement is placed in this form and held in placeduring the concreting operation. After the concrete has hardened, the forms are removed. As the forms are removed, props of shores are installed to support the weight of the concrete until it has reached sufficient strength to support the loads by itself.The designer must proportion a concrete member for adequate strength to resist the loads and adequate stiffness to prevent excessive deflections. In beam must be proportioned so that it can be constructed. For example, the reinforcement must be detailed so that it can be assembled in the field, and since the concrete is placed in the form after the reinforcement is in place, the concrete must be able to flow around, between, and past the reinforcement to fill all parts of the form completely.The choice of whether a structure should be built of concrete, steel, masonry, or timber depends on the availability of materials and on a number of value decisions. The choice of structural system is made by the architect of engineer early in the design, based on the following considerations:1. Economy. Frequently, the foremost consideration is the overall const of the structure. This is, of course, a function of the costs of the materials and the labor necessary to erect them. Frequently, however, the overall cost is affected as much or more by the overall construction time since the contractor and owner must borrow or otherwise allocate money to carry out the construction and will not receive a return on this investment until the building is ready for occupancy. In a typical large apartment of commercial project, the cost of construction financing will be a significant fraction of the total cost. As a result, financial savings due to rapid construction may more than offset increased material costs. For this reason, any measures the designer can take to standardize the design and forming will generally pay off in reduced overall costs.In many cases the long-term economy of the structure may be more important than the first cost. As a result, maintenance and durability are important consideration.2. Suitability of material for architectural and structural function.A reinforced concrete system frequently allows the designer to combine the architectural and structural functions. Concrete has the advantage that it is placed in a plastic condition and is given the desired shapeand texture by means of the forms and the finishing techniques. This allows such elements ad flat plates or other types of slabs to serve as load-bearing elements while providing the finished floor and / or ceiling surfaces. Similarly, reinforced concrete walls can provide architecturally attractive surfaces in addition to having the ability to resist gravity, wind, or seismic loads. Finally, the choice of size of shape is governed by the designer and not by the availability of standard manufactured members.3. Fire resistance. The structure in a building must withstand the effects of a fire and remain standing while the building is evacuated and the fire is extinguished. A concrete building inherently has a 1- to 3-hour fire rating without special fireproofing or other details. Structural steel or timber buildings must be fireproofed to attain similar fire ratings.4. Low maintenance.Concrete members inherently require less maintenance than do structural steel or timber members. This is particularly true if dense, air-entrained concrete has been used for surfaces exposed to the atmosphere, and if care has been taken in the design to provide adequate drainage off and away from the structure. Special precautions must be taken for concrete exposed to salts such as deicing chemicals.5. Availability of materials. Sand, gravel, cement, and concrete mixing facilities are very widely available, and reinforcing steel can be transported to most job sites more easily than can structural steel. As a result, reinforced concrete is frequently used in remote areas.On the other hand, there are a number of factors that may cause one to select a material other than reinforced concrete. These include:1. Low tensile strength.The tensile strength concrete is much lower than its compressive strength ( about 1/10 ), and hence concrete is subject to cracking. In structural uses this is overcome by using reinforcement to carry tensile forces and limit crack widths to within acceptable values. Unless care is taken in design and construction, however, these cracks may be unsightly or may allow penetration of water. When this occurs, water or chemicals such as road deicing salts may cause deterioration or staining of the concrete. Special design details are required in such cases. In the case of water-retaining structures, special details and /of prestressing are required to prevent leakage.2. Forms and shoring. The construction of a cast-in-place structure involves three steps not encountered in the construction of steel or timber structures. These are ( a ) the construction of the forms, ( b ) the removal of these forms, and (c) propping or shoring the new concrete to support its weight until its strength is adequate. Each of these steps involves labor and / or materials, which are not necessary with other forms of construction.3. Relatively low strength per unit of weight for volume.The compressive strength of concrete is roughly 5 to 10% that of steel, while its unit density is roughly 30% that of steel. As a result, a concrete structure requires a larger volume and a greater weight of material than does a comparable steel structure. As a result, long-span structures are often built from steel.4. Time-dependent volume changes. Both concrete and steel undergo-approximately the same amount of thermal expansion and contraction. Because there is less mass of steel to be heated or cooled, and because steel is a better concrete, a steel structure is generally affected by temperature changes to a greater extent than is a concrete structure. On the other hand, concrete undergoes frying shrinkage, which, if restrained, may cause deflections or cracking. Furthermore, deflections will tend to increase with time, possibly doubling, due to creep of the concrete under sustained loads.In almost every branch of civil engineering and architecture extensive use is made of reinforced concrete for structures and foundations. Engineers and architects requires basic knowledge of reinforced concrete design throughout their professional careers. Much of this text is directly concerned with the behavior and proportioning of components that make up typical reinforced concrete structures-beams, columns, and slabs. Once the behavior of these individual elements is understood, the designer will have the background to analyze and design a wide range of complex structures, such as foundations, buildings, and bridges, composed of these elements.Since reinforced concrete is a no homogeneous material that creeps, shrinks, and cracks, its stresses cannot be accurately predicted by the traditional equations derived in a course in strength of materials forhomogeneous elastic materials. Much of reinforced concrete design in therefore empirical, i.e., design equations and design methods are based on experimental and time-proved results instead of being derived exclusively from theoretical formulations.A thorough understanding of the behavior of reinforced concrete will allow the designer to convert an otherwise brittle material into tough ductile structural elements and thereby take advantage of concrete’s desirable characteristics, its high compressive strength, its fire resistance, and its durability.Concrete, a stone like material, is made by mixing cement, water, fine aggregate ( often sand ), coarse aggregate, and frequently other additives ( that modify properties ) into a workable mixture. In its unhardened or plastic state, concrete can be placed in forms to produce a large variety of structural elements. Although the hardened concrete by itself, i.e., without any reinforcement, is strong in compression, it lacks tensile strength and therefore cracks easily. Because unreinforced concrete is brittle, it cannot undergo large deformations under load and fails suddenly-without warning. The addition fo steel reinforcement to the concrete reduces the negative effects of its two principal inherent weaknesses, its susceptibility to cracking and its brittleness. When the reinforcement is strongly bonded to the concrete, a strong, stiff, and ductile construction material is produced. This material, called reinforced concrete, is used extensively to construct foundations, structural frames, storage takes, shell roofs, highways, walls, dams, canals, and innumerable other structures and building products. Two other characteristics of concrete that are present even when concrete is reinforced are shrinkage and creep, but the negative effects of these properties can be mitigated by careful design.A code is a set technical specifications and standards that control important details of design and construction. The purpose of codes it produce structures so that the public will be protected from poor of inadequate and construction.Two types f coeds exist. One type, called a structural code, is originated and controlled by specialists who are concerned with the proper use of a specific material or who are involved with the safe design of a particular class of structures.The second type of code, called a building code, is established to cover construction in a given region, often a city or a state. The objective of a building code is also to protect the public by accounting for the influence of the local environmental conditions on construction. For example, local authorities may specify additional provisions to account for such regional conditions as earthquake, heavy snow, or tornados. National structural codes genrally are incorporated into local building codes.The American Concrete Institute ( ACI ) Building Code covering the design of reinforced concrete buildings. It contains provisions covering all aspects of reinforced concrete manufacture, design, and construction. It includes specifications on quality of materials, details on mixing and placing concrete, design assumptions for the analysis of continuous structures, and equations for proportioning members for design forces.All structures must be proportioned so they will not fail or deform excessively under any possible condition of service. Therefore it is important that an engineer use great care in anticipating all the probable loads to which a structure will be subjected during its lifetime.Although the design of most members is controlled typically by dead and live load acting simultaneously, consideration must also be given to the forces produced by wind, impact, shrinkage, temperature change, creep and support settlements, earthquake, and so forth.The load associated with the weight of the structure itself and its permanent components is called the dead load. The dead load of concrete members, which is substantial, should never be neglected in design computations. The exact magnitude of the dead load is not known accurately until members have been sized. Since some figure for the dead load must be used in computations to size the members, its magnitude must be estimated at first. After a structure has been analyzed, the members sized, and architectural details completed, the dead load can be computed more accurately. If the computed dead load is approximately equal to the initial estimate of its value ( or slightly less ), the design is complete, but if a significant difference exists between the computed and estimated values of dead weight, the computations should be revised using an improved value of dead load. An accurate estimate of dead load is particularly important when spans are long, say over 75 ft ( 22.9 m ),because dead load constitutes a major portion of the design load.Live loads associated with building use are specific items of equipment and occupants in a certain area of a building, building codes specify values of uniform live for which members are to be designed.After the structure has been sized for vertical load, it is checked for wind in combination with dead and live load as specified in the code. Wind loads do not usually control the size of members in building less than 16 to 18 stories, but for tall buildings wind loads become significant and cause large forces to develop in the structures. Under these conditions economy can be achieved only by selecting a structural system that is able to transfer horizontal loads into the ground efficiently.钢筋混凝土在每一个国家，混凝土及钢筋混凝土都被用来作为建筑材料。

土木工程毕业设计外文翻译---建筑结构整体设计-建筑创作和综合技术

Create and comprehensive technology in the structure globaldesign of the buildingThe 21st century will be the era that many kinds of disciplines technology coexists , it will form the enormous motive force of promoting the development of building , the building is more and more important too in global design, the architect must seize the opportunity , give full play to the architect's leading role, preside over every building engineering design well. Building there is the global design concept not new of architectural design,characteristic of it for in an all-round way each element not correlated with building- there aren't external environment condition, building , technical equipment,etc. work in coordination with, and create the premium building with the comprehensive new technology to combine together.The premium building is created, must consider sustainable development , namely future requirement , in other words, how save natural resources as much as possible, how about protect the environment that the mankind depends on for existence, how construct through high-quality between architectural design and building, in order to reduce building equipment use quantity and reduce whole expenses of project.The comprehensive new technology is to give full play to the technological specialty of every discipline , create and use the new technology, and with outside space , dimension of the building , working in coordination with in an all-round way the building component, thus reduce equipment investment and operate the expenses.Each success , building of engineering construction condense collective intelligence and strength; It is intelligence and expectation that an architect pays that the building is created; The engineering design of the building is that architecture , structure , equipment speciality compose hardships and strength happenning; It is the diligent and sweat paid in design and operation , installation , management that the construction work is built up .The initial stage of the 1990s, our understanding that the concept of global design is a bit elementary , conscientious to with making some jobs in engineeringdesign unconsciously , make some harvest. This text Hangzhou city industrial and commercial bank financial comprehensive building and Hangzhou city Bank of Communications financial building two building , group of " scientific and technological progress second prize " speak of from person who obtain emphatically, expound the fact global design - comprehensive technology that building create its , for reach global design outstanding architect in two engineering design, have served as the creator and persons who cooperate while every stage design and even building are built completely.Two projects come into operation for more than 4 years formally , run and coordinate , good wholly , reach the anticipated result, accepted and appreciated by the masses, obtain various kinds of honor .outstanding to design award , progress prize in science and technology , project quality bonus , local top ten view , best model image award ,etc., the ones that do not give to the architect and engineers without one are gratified and proud. The building is created Emphasizing the era for global design of the building, the architects' creation idea and design method should be broken through to some extent, creation inspirations is it set up in analysis , building of global design , synthesize more to burst out and at the foundation that appraise, learn and improve the integration capability exactly designed in building , possess the new knowledge system and thinking method , merge multi-disciplinary technology. We have used the new design idea in above-mentioned projects, have emphasized the globality created in building .Is it is it act as so as to explain to conceive to create two design overview and building of construction work these now.1) The financial comprehensive building of industrial and commercial bank of HangZhou, belong to the comprehensive building, with the whole construction area of 39,000 square meters, main building total height 84, 22, skirt 4 of room, some 6 storeys, 2 storeys of basements.Design overall thinking break through of our country bank building traditional design mode - seal , deep and serious , stern , form first-class function, create of multi-functional type , the style of opening , architecture integrated with the mode of the international commercial bank.The model of the building is free and easy, opened, physique was made up by the hyperboloid, the main building presented " the curved surface surrounded southwards ", skirt room presents " the curved surface surrounded northwards ", the two surround but become intension of " gathering the treasure ".Building flourishing upwards, elevation is it adopt large area solid granite wall to design, the belt aluminium alloy curtain wall of the large area and some glass curtain walls, and interweave the three into powerful and vigorous whole , chase through model and entity wall layer bring together , form concise , tall and straight , upward tendency of working up successively, have distinct and unique distinctions.Building level and indoor space are designed into a multi-functional type and style of opening, opening, negotiate , the official working , meeting , receiving , be healthy and blissful , visit combining together. Spacious and bright two storeys open in the hall unifiedly in the Italian marble pale yellow tone , in addition, the escalator , fountain , light set off, make the space seem very magnificent , graceful and sincere. Intelligent computer network center, getting open and intelligent to handle official business space and all related house distribute in all floor reasonably. Top floor round visit layer, lift all of Room visit layer , can have a panoramic view of the scenery of the West Lake , fully enjoy the warmth of the nature. 2) The financial building of Bank of Communications of Hangzhou, belong to the purely financial office block, with the whole construction area of 19,000 square meters, the total height of the building is 39.9 meters, 13 storeys on the ground, the 2nd Floor. Live in building degree high than it around location , designer have unique architectural appearance of style architectural design this specially, its elevation is designed into a new classical form , the building base adopts the rough granite, show rich capability , top is it burn granite and verticality bar and some form aluminum windows make up as the veneer to adopt, represent the building noble and refined , serious personality of the bank.While creating in above-mentioned two items, besides portraying the shape of the building and indoor space and outside environment minister and blending meticulously, in order to achieve the outstanding purpose of global design of the building , the architect , still according to the region and project characteristic, putforward the following requirement to every speciality:(1) Control the total height of the building strictly;(2) It favorable to the intelligent comfortable height of clearances to create;(3) Meet the floor area of owner's demand;(4)Protect the environment , save the energy , reduce and make the investment;(5) Design meticulously, use and popularize the new technology;(6) Cooperate closely in every speciality, optimization design. Comprehensive technologyThe building should have strong vitality, there must be sustainable development space, there should be abundant intension and comprehensive new technology. Among above-mentioned construction work , have popularized and used the intelligent technology of the building , has not glued and formed the flat roof beam of prestressing force - dull and stereotyped structure technology and flat roof beam structure technology, baseplate temperature mix hole , technology of muscle and base of basement enclose new technology of protecting, computer control STL ice hold cold air conditioner technology, compounding type keeps warm and insulates against heat the technology of the wall , such new technologies as the sectional electricity distribution room ,etc., give architecture global design to add the new vitality of note undoubtedly.1, the intelligent technology of the buildingIn initial stage of the 1990s, the intelligent building was introduced from foreign countries to China only as a kind of concept , computer network standard is it soon , make information communication skeleton of intelligent building to pursue in the world- comprehensive wiring system becomes a kind of trend because of 10BASE-T. In order to make the bank building adapt to the development of the times, the designer does one's utmost to recommend and design the comprehensive wiring system with the leading eyes , this may well be termed the first modernized building which adopted this technical design at that time.(1) Comprehensive wiring system one communication transmission network, it make between speech and data communication apparatus , exchange equipment andother administrative systems link to each other, make the equipment and outside communication network link to each other too. It include external telecommunication connection piece and inside information speech all cable and relevant wiring position of data terminal of workspace of network. The comprehensive wiring system adopts the products of American AT&T Corp.. Connected up the subsystem among the subsystem , management subsystem , arterial subsystem and equipment to make up by workspace subsystem , level.(2) Automated systems of security personnel The monitoring systems of security personnel of the building divide into the public place and control and control two pieces of system equipment with the national treasury special-purposly synthetically.The special-purpose monitoring systems of security personnel of national treasury are in the national treasury , manage the storehouse on behalf of another , transporting the paper money garage to control strictly, the track record that personnel come in and go out, have and shake the warning sensor to every wall of national treasury , the camera, infrared microwave detector in every relevant rooms, set up the automation of controlling to control.In order to realize building intellectuality, the architect has finished complete indoor environment design, has created the comfortable , high-efficient working environment , having opened up the room internal and external recreation space not of uniform size, namely the green one hits the front yard and roofing, have offered the world had a rest and regulated to people working before automation is equipped all day , hang a design adopt the special building to construct the node in concrete ground , wall at the same time.2, has not glued and formed the flat roof beam of prestressing force- dull and stereotyped structure technology and flat roof beam structure technology In order to meet the requirement with high assurance that the architect puts forward , try to reduce the height of structure component in structure speciality, did not glue and form the flat roof beam of prestressing force concrete - dull and stereotyped structure technology and flat roof beam structure technology after adopting.(1) Adopt prestressing force concrete roof beam board structure save than ordinary roof beam board concrete consumption 15%, steel consumption saves 27%, the roof beam reduces 300mm high.(2) Adopt flat roof beam structure save concrete about 10% consumption than ordinary roof beam board, steel consumption saves 6.6%, the roof beam reduces 200mm high.Under building total situation that height does not change , adopt above-mentioned structure can make the whole building increase floor area of a layer , have good economic benefits and social benefit.3, the temperature of the baseplate matches muscle technologyIn basement design , is it is it is it after calculating , take the perimeter to keep the construction technology measure warm to split to resist to go on to baseplate, arrange temperature stress reinforcing bar the middle cancelling , dispose 2 row receives the strength reinforcing bar up and down only, this has not only save the fabrication cost of the project but also met the basement baseplate impervious and resisting the requirement that splits.4, the foundation of the basement encloses and protects the new technology of design and operationAdopt two technological measures in enclosing and protecting a design:(1) Cantilever is it is it hole strength is it adopt form strengthen and mix muscle technology to design to protect to enclose, save the steel and invite 60t, it invests about 280,000 to save.(2) Is it is it protect of of elevation and keep roof beam technology to enclose , is it protect long to reduce 1.5m to enclose all to reduce, keep roof beam mark level on natural ground 1.5m , is it is it protect of lateral pressure receive strength some height to enclose to change, saving 137.9 cubic meters of concrete, steel 16.08t, reduces and invests 304,000 yuan directly through calculating.5, ice hold cold air conditioner technologyIce hold cold air conditioner technology belong to new technology still in our country , it heavy advantage that the electricity moves the peak and operates theexpenses sparingly most. In design, is it ice mode adopt some (weight ) hold mode of icing , is it ice refrigeration to be plane utilization ratio high to hold partly to hold, hold cold capacity little , refrigeration plane capacity 30%-45% little than routine air conditioner equipment, one economic effective operational mode.Hold the implementation of the technology of the cold air conditioner in order to cooperate with the ice , has used intelligent technology, having adopted the computer to control in holding and icing the air conditioner system, the main task has five following respects:(1) According to the demand for user's cold load , according to the characteristic of the structure of the electric rate , set up the ice and hold the best operation way of the cold system automatically, reduce the operation expenses of the whole system;(2) Fully utilize and hold the capacity of the cold device, should try one's best to use up all the cold quantity held basically on the same day;(3) Automatic operation state of detection system, ensure ice hold cold system capital equipment normal , safe operation;(4) Automatic record parameter that system operate, display system operate flow chart and type systematic operation parameter report form;(5) Predict future cooling load, confirm the future optimization operation scheme.Ice hold cold air conditioner system test run for some time, indicate control system to be steady , reliable , easy to operate, the system operates the energy-conserving result remarkably.6, the compounding type keeps in the wall warm and insulates against heat To the area of Hangzhou , want heating , climate characteristic of lowering the temperature in summer in winter, is it protect building this structural design person who compound is it insulate against heat the wall to keep warm to enclose specially, namely: Fit up , keep warm , insulate against heat the three not to equal to the body , realize building energy-conservation better.Person who compound is it insulate against heat wall to combine elevation model characteristic , design aluminium board elevation renovation material to keepwarm, its structure is: Fill out and build hollow brick in the frame structure, do to hang the American Fluorine carbon coating inferior mere aluminium board outside the hollow brick wall.Aluminium board spoke hot to have high-efficient adiabatic performance to the sun, under the same hot function of solar radiation, because the nature , color of the surface material are different from coarse degree, whether can absorb heat have great difference very , between surface and solar radiation hot absorption system (α ) and material radiation system (Cλ ) is it say to come beyond the difference this. Adopt α and Cλ value little surface material have remarkable result , board α、Cλ value little aluminium have, its α =0.26, Cλ =0.4, light gray face brick α =0.56, Cλ =4.3.Aluminium board for is it hang with having layer under air by hollow brick to do, because aluminium board is it have better radiation transfer to hot terms to put in layer among the atmosphere and air, this structure is playing high-efficient adiabatic function on indoor heating too in winter, so, no matter or can well realize building energy-conservation in winter in summer.7, popularize the technology of sectional electricity distribution roomConsider one layer paves Taxi " gold " value , the total distribution of the building locates the east, set up voltage transformer and low-voltage distribution in the same room in first try in the design, make up sectional electricity distribution room , save transformer substation area greatly , adopt layer assign up and down, mixing the switchyard system entirely after building up and putting into operation, the function is clear , the overall arrangement compactness is rational , the systematic dispatcher is flexible . The technology have to go to to use and already become the model extensively of the design afterwards.ConclusionThe whole mode designed of the building synthetically can raise the adaptability of the building , it will be the inevitable trend , environmental consciousness and awareness of saving energy especially after strengthening are even more important. Developing with the economy , science and technology constantly in our country, more advanced technology and scientific and technical result will be applied to thebuilding , believe firmly that in the near future , more outstanding building global design will appear on the building stage of our country. We will be summarizing, progressing constantly constantly, this is that history gives the great responsibility of architect and engineer.汉语翻译建筑结构整体设计-建筑创作和综合技术21世纪将是多种学科技术并存的时代，它必将形成推动建筑发展的巨大动力，建筑结构整体设计也就越来越重要，建筑师必须把握时机，充分发挥建筑师的主导作用，主持好各项建筑工程设计。

土木工程毕业论文中英文翻译

一、科技资料原文:Structural Systems to resist lateral loadsCommonly Used structural SystemsWith loads measured in tens of thousands kips，there is little room in the design of high—rise buildings for excessively complex thoughts。

Indeed，the better high—rise buildings carry the universal traits of simplicity of thought and clarity of expression.It does not follow that there is no room for grand thoughts。

Indeed, it is with such grand thoughts that the new family of high—rise buildings has evolved. Perhaps more important，the new concepts of but a few years ago have become commonplace in today’ s technology。

Omitting some concepts that are related strictly to the materials of construction, the most commonly used structural systems used in high-rise buildings can be categorized as follows:1.Moment—resisting frames。

2.Braced frames, including eccentrically braced frames.3.Shear walls，including steel plate shear walls。

1、下载文档前请自行甄别文档内容的完整性，平台不提供额外的编辑、内容补充、找答案等附加服务。
2、"仅部分预览"的文档,不可在线预览部分如存在完整性等问题,可反馈申请退款(可完整预览的文档不适用该条件!)。
3、如文档侵犯您的权益，请联系客服反馈,我们会尽快为您处理(人工客服工作时间：9:00-18:30)。

附录1 外文翻译原文3.2Elastic models3.2.1AnisotropyAn isotropic material has the same properties in all directions —we cannot dis-tinguish any one direction from any other. Samples taken out of the ground with any orientation would behave identically. However, we know that soils have been deposited in some way —for example, sedimentary soils will know about the vertical direction of gravitational deposition. There may in addition be seasonal variations in the rate of deposition so that the soil contains more or less marked layers of slightly different grain size and/or plasticity. The scale of layering may be suffciently small that we do not wish to try to distinguish separate materials, but the layering together with the directional deposition may nevertheless be suffcient to modify the properies of the soil in different directions —in other words to cause it to be anisotropic.We can write the stiffness relationship between elastic strain increment e δε and stress increment δσ compactly ase D δεδσ= )36.3(where D is the stiffness matrix and hence 1-D is the compliance matrix. For a completely general anisotropic elastic material⎪⎪⎪⎪⎪⎪⎪⎪⎭⎫⎝⎛=-u t r o k f t s q n j e r q p m i d o n m l h c k j i hg b f e d c b a D 1)37.3( whereeachlettera,b,... is,inprinciple,anindependentelasticpropertyandthe necessarysymmetry of the sti?ness matrix for the elastic material has reduced the maximum number of independent properties to 21. As soon as there are material symmetries then the number of independent elastic properties falls (Crampin, 1981).For example, for monoclinic symmetry (z symmetry plane) the compliance matrix has the form:⎪⎪⎪⎪⎪⎪⎪⎪⎭⎫ ⎝⎛=-m igd l k k j i h f c g fe b d c b a D 0000000000000001)38.3(and has thirteen elastic constants. Orthorhombic symmetry (distinct x, y and z symmetry planes) gives nine constants:⎪⎪⎪⎪⎪⎪⎪⎪⎭⎫⎝⎛=-i h g f e c e d b c ba D 0000000000000000000000001)39.3( whereas cubic symmetry (identical x, y and z symmetry planes, together with planes joining opposite sides of a cube) gives only three constants:⎪⎪⎪⎪⎪⎪⎪⎪⎭⎫⎝⎛=-c c c a b b b a b b b a D 0000000000000000000000001)40.3(Figure 3.9: Independent modes of shearing for cross-anisotropic materialIf we add the further requirement that )(2b a c -=and set E a /1= andE v b /-=，then we recover the isotropic elastic compliance matrix of (3.1).Though it is obviously convenient if geotechnical materials have certain fabric symmetries which confer a reduction in the number of independent elastic properties,it has to be expected that in general materials which have been pushed around by tectonic forces, by ice, or by man will not possess any of these symmetries and, insofar as they have a domain of elastic response, we should expect to require the full 21 independent elastic properties. If we choose to model such materials as isotropic elastic or anisotropic elastic with certain restricting symmetries then we have to recognise that these are modelling decisions of which the soil or rock may be unaware.However, many soils are deposited over areas of large lateral extent and symmetry of deposition is essentially vertical. All horizontal directions look the same but horizontal sti?ness is expected to be di?erent from vertical stiffness. The form of the compliance matrix is now:⎪⎪⎪⎪⎪⎪⎪⎪⎭⎫ ⎝⎛=-f e e d c c c a b c b a D000000000000000000001)41.3( and we can write ：:/)1(2)(2/1,/1,/,/,/1h hh vh v v vh h hh h E v b a f G e E d E v c E v b E a +=-===-=-==和=-1D ()⎪⎪⎪⎪⎪⎪⎪⎪⎭⎫⎝⎛+------h hh hv hv h vvh v vh v vh h h hh v vh h hh hE v G G E E v Ev E v E E v E v E v E /12000/1000000/1000000/1//000//1/000///1 )42.3(This is described as transverse isotropy or cross anisotropy with hexagonal symmetry. There are 5 independent elastic properties: v E and h E are Young’s moduli forunconfined compression in the vertical and horizontal directions respectively; h v G is the shear modulus for shearing in a vertical plane (Fig 3.9a).Poisson’s ratios h h V and h v V relate to the lateral strains that occur in the horizontal direction orthogonal to a horizontal direction of compression and a vertical direction of compression respectively (Fig 3.9c, b).Testing of cross anisotropic soils in a triaxial apparatus with their axes of anisotropy aligned with the axes of the apparatus does not give us any possibility todiscover h v G ()E /1=，since this would require controlled application of shear stresses to vertical and horizontal surfaces of the sample —and attendant rotation of principal axes. In fact we are able only to determine 3 of the 5 elastic properties. If we write (3.42) for radial and axial stresses and strains for a sample with its vertical axis of symmetry of anisotropy aligned with the axis of the triaxial apparatus, we find that ：()⎪⎪⎭⎫⎝⎛⎪⎪⎭⎫⎝⎛---=⎪⎪⎭⎫ ⎝⎛''/1//2/1r a v hh v vh v vh v r a E v E v E v E δσδσδεδε )43.3( The compliance matrix is not symmetric because, in the context of the triaxial test, the strain increment and stress quantities are not properly work conjugate. We deduce that while we can separately determine v E and h v V the only other elastic property that we can discover is the composite stiffness )1/(hh h V E -.We are not able to separate h E and h h V (Lings et al., 2000).On the other hand, Graham and Houlsby (1983) have proposed a special form of(3.41) or (3.42) which uses only 3 elastic properties but forces certain interdependencies among the 5 elastic properties for this cross anisotropicmaterial.()()()⎪⎪⎪⎪⎪⎪⎪⎪⎭⎫⎝⎛+++------⨯=**********-222221/1200000/12000000/120000001//000//1/000///11αααααααααααv v v v v v v v v E D)44.3(This is written in terms of a Young ’s modulus v E E =*,the Young ’s modulus for loading in the vertical direction, a Poisson ’s ratio hh V V =*,together with a third parameter α. The ratio of stiffness in horizontal and vertical directions is2/α=v h E E and other linkages are forced ：)1(2//;/**+===v E G G v v hh hv hh vh ααα.For our triaxial stress and strain quantities, the compliance matrix becomes:⎪⎪⎭⎫⎝⎛⎪⎪⎭⎫ ⎝⎛--=⎪⎪⎭⎫ ⎝⎛**q p K JJ G q p δδδεδε'3det 1 )45.3(Figure 3.10: Effect of cross-anisotropy on direction of undrained effective stress pathwhere23det J G K -=** )46.3(and the stiffness matrix is⎪⎪⎭⎫⎝⎛⎪⎪⎭⎫ ⎝⎛=⎪⎪⎭⎫ ⎝⎛**q p G JJ K q p δεδεδδ3' )47.3(where()()******-+++-=v v v v EK 21192412αα )48.3( ()()******-++--=vv v v E G 21164222αα )49.3( ()()******-+++-=v v v v EJ 211312αα )50.3( The stiffness and compliance matrices (written in terms of correctly chosen work conjugate strain increment and stress quantities) are still symmetric —the material is still elastic —but the non-zero off-diagonal terms tell us that there is now coupling between volumetric and distortional effects. There will be volumetric strain when we apply purely distortional stress, 0'=p δ,distortional strain during purely isotropiccompression, 0=q δ,and there will be change in mean effective stress in undrained tests,0=p δε.In fact the slope of the effective stress path in an undrained test is, from (3.45),()()224223123'ααααδδ+---+-==*****v v v v G J q p )51.3( From our definition of pore pressure parameter a (§2.6.2) we find*-=-=GJq p 3'δδα )52.3(Figure 3.11: Relationship between anisotropy parameter α and pore pressureparameter a for different values of Poisson ’s ratio ν*.which will, in the presence of anisotropy, not be zero.A first inspection of (3.51) merely suggests that there are limits on the pore pressure parameter of a = 2/3 and a = -1/3 for α very large （h E ＞＞v E ）and α verysmall （v E ＞＞h E ）repectively (Fig 3.10), which in turn imply effective stress paths with constant axial effective stress and constant radial effective stress respectively. The link between a and α is actually slightly more subtle.In fact,for 0≠*v the relationship is not actually monotonic and the effective stress path direction overshoots the apparent limits (Fig 3.11). The deduction of a valueof α(and hence 2/α=v h E E ) from a is not very reliable when a is around -1/3 or 2/3 (recallthe data presented in Figs 2.51 and 2.49, §2.5.4). For 5.0=*v ,[])1(3/)21(αα-+-=a or)23)(31(-+=a a α.These relationships satisfy the expected limits for 0=α and ∞=αbutthere are singularities in the inversion of (3.51) for 1=α and 5.0=*v .3.2.2NonlinearityWe will probably expect that the dominant source of nonlinearity of stress:strain response will come from material plasticity —and we will go on to develop elasticplastic constitutive models in the next section. However, we also have an expectation that some of the truly elastic properties of soils will vary with stress level and this can be seen as a source of elastic nonlinearity. Our thoughts about elastic materials as conservative materials —the term ‘hyperelasticity ’ is used to describe such materials —might make us a little cautious about plucking from the air arbitrary empirical functions for variation of moduli with stresses. For example, if we were to suppose that the bulk modulus of the soil varied with mean effective stress but that Poisson ’s ratio (and hence the ratio of shear modulus to bulk modulus) were constant then we would find that in a closed stress cycle such as that shown in Fig 3.12 energy would be created (or lost) creating a perpetual motion machine in violation of the first law of thermodynamics —this would not be a conservative system. We need to find a strain energy (3.7) or complementary energy density (3.11) function which can be differentiated to give acceptable variation of moduli with stresses.Figure 3.12: Cycle of stress changes which should give zero energy generated ordissipated for conservative materialSuch a complementary energy function can be deduced from the nonlinear elastic model described by Boyce (1980):()⎪⎪⎭⎫ ⎝⎛⎥⎦⎤⎢⎣⎡++=+211'6111'p q G K n p V n )53.3( When 1K and 1G are reference values of bulk modulus and shear modulus and n is anonlinearity parameter. The compliance matrix can then be deduced by differentiation ：()()⎪⎪⎭⎫⎝⎛⎪⎪⎪⎪⎭⎫⎝⎛----+++=⎪⎪⎭⎫ ⎝⎛+q p G G n G n G n n K np n q p δδηηηδεδε'313131621'1112111 )54.3( Where '/p q =η.There is again (as for the anisotropic model) coupling between volumetric and distortional effects. The stiffnesses are broadly proportional to n p -1'. Because the compliances are now varying with stress ratio η the effective stress path implied for an undrained (purely distortional) loading is no longer straight. In fact, for a reference state 0,'0===q p p ηthe effective stress path is()np p 201''βη-= )55.3( where 116/)1(G K n -=β。

外文翻译(土木工程专业毕业设计)

土木工程专业毕业设计- 外文翻译

土木工程专业毕业设计外文文献及翻译

土木工程毕业外文翻译

土木工程毕业设计--外文翻译

土木工程专业毕业设计外文翻译

土木工程毕业设计外文翻译

土木工程--毕业设计外文翻译(原文+翻译)

土木工程毕业设计 外文翻译

土木专业毕业设计外文翻译英文文献

土木工程建筑工程毕业设计外文翻译

土木工程建筑工程毕业设计外文翻译

土木工程毕业设计外文翻译最终中英文

土木工程外文翻译

(完整版)土木工程毕业设计外文文献翻译

土木工程毕业设计外文翻译---建筑结构整体设计-建筑创作和综合技术

土木工程毕业论文中英文翻译

土木工程毕业设计外文翻译