土木工程中英文翻译

from：journal of Constructional Steel Research.V olume 59,Number 1,January 2003 Cyclic behavior of steel moment frameconnections under varying axial load and lateral displacements Abstract: This paper discusses the cyclic behavior of four steel moment connections tested under variable axial load and lateral displacements. The beam specim- ens consisted of a reduced beam section, wing plates and longitudinal stiffeners. The test specimens were subjected to varying axial forces and lateral displace- ments to simulate the effects on beams in a Coupled-Girder Moment-Resisting Framing system under lateral loading. The test results showed that the specim- ens responded in a ductile manner since the plastic rotations exceeded 0.03 rad without significant drop in the lateral capacity. The presence of the longitudin- al stiffener assisted in transferring the axial forces and delayed the formation of web local buckling.1. IntroductionAimed at evaluating the structural performance of reduced-beam section(RBS) connections under alternated axial loading and lateral displacement, four full-scale specimens were tested. These tests were intended to assess the performance of the moment connection design for the Moscone Center Exp- ansion under the Design Basis Earthquake (DBE) and the Maximum Considered Earthquake (MCE). Previous research conducted on RBS moment connections [1,2] showed that connections with RBS profiles can achieve rotations in excess of 0.03 rad. However, doubts have been cast on the quality of the seismic performance of these connections under combined axial and lateral loading.The Moscone Center Expansion is athree-story, 71,814 m2 (773,000 ft2) structurewith steel moment frames as its primary lateralforce-resisting system. A three dimensionalperspective illustration is shown in Fig. 1. Theoverall height of the building, at the highest point of the exhibition roof, is approxima- tely 35.36 m (116ft) above ground level. The ceiling height at the exhibition hall is 8.23 m (27 ft) , and the typical floor-to-floor height in the building is 11.43 m (37.5 ft). The building was designed as type I according to the requi- rements of the 1997 Uniform Building Code.The framing system consists of four moment frames in the East–West direct- ion, one on either side of the stair towers, and four frames in the North–Southdirection, one on either side of the stair and elevator cores in the east end and two at the west end of the structure [4]. Because of the story height, the con- cept of the Coupled-Girder Moment-Resisting Framing System (CGMRFS) was utilized.By coupling the girders, the lateral load-resisting behavior of the moment framing system changes to one where structural overturning moments are resisted partially by an axial compression–tension couple across the girder system, rather than only by the individual flexural action of the girders. As a result, a stiffer lateral load resisting system is achieved. The vertical element that connects the girders is referred to as a coupling link. Coupling links are analogous to and serve the same structural role as link beams in eccentrically braced frames. Coupling links are generally quite short, having a large shear- to-moment ratio.Under earthquake-type loading, the CGMRFS subjects its girders to wariab- ble axial forces in addition to their end moments. The axial forces in theFig. 1. Moscone Center Expansion Project in San Francisco, CAgirders result from the accumulated shear in the link.2. Analytical model of CGMRFNonlinear static pushover analysis wasconducted on a typical one-bay model of theCGMRF. Fig. 2 shows the dimensions andthe various sections of the model. The linkflange plates were 28.5 mm ⋅ 254 mm (1 1/8 in ⋅ 10 in) and the web plate was 9.5 mm ⋅ 476mm (3 /8 in ⋅ 18 3/4 in). The SAP 2000computer program was utilized in thepushover analysis [5]. The frame was characterized as fully restrained(FR). FR moment frames are those frames for 1170which no more than 5% of the lateral deflections arise from connection deformation [6]. The 5% value refers only to deflection due to beam–column deformation and not to frame deflections that resultfrom column panel zone deformation [6, 9].The analysis was performed using an expected valueof the yield stress and ultimate strength. These valueswere equal to 372 MPa (54 ksi) and 518 MPa (75 ksi),respectively. The plastic hinges’ load–deformationbehavior was approximated by the generalized curvesuggested by NEHRP Guidelines for the SeismicRehabilitation of Buildings [6] as shown in.Fig. 3. △ywas calcu- lated based on Eqs. (5.1) and (5.2) from[6], as follows:P–M hinge load–deformation model points C, D and E are based on Table 5.4 from [6] for △y was taken as 0.01 rad per Note 3 in [6], Table 5.8. Shear hinge load- load–deformation model points C, D and E are based on Table 5.8 [6], Link Beam, Item a. A strain hardening slope between points B and C of 3% of the elastic slope was assumed for both models.The following relationship was used to account for moment–axial load interaction [6]:where MCE is the expected moment strength, ZRBS is the RBS plastic section modulus (in3), is the expected yield strength of the material (ksi), P is the axial force in the girder (kips) and is the expected axial yield force of the RBS, equal to (kips). The ultimate flexural capacities of the beam and the link of the model are shown in Table 1.Fig. 4 shows qualitatively the distribution of the bending moment, shear force, and axial force in the CGMRF under lateral load. The shear and axial force in the beams are less significant to the response of the beams as compared with the bending moment, although they must be considered in design. The qualita- tive distribution of internal forces illustrated in Fig. 5 is fundamentally the same for both elastic and inelastic ranges of behavior. The specific values of the internal forces will change as elements of the frame yield and internal for- ces are redistributed. The basic patterns illustrated in Fig. 5, however, remain the same.Inelastic static pushover analysis was carried out by applying monotonically increasing lateral displacements, at the top of both columns, as shown in Fig. 6. After the four RBS have yielded simultaneously, a uniform yielding in the web and at the ends of the flanges of the vertical link will form. This is the yield mechanism for the frame , with plastic hinges also forming at the base of the columns if they are fixed. The base shear versus drift angle of the model is shown in Fig. 7 . The sequence ofinelastic activity in the frame is shown on the figure. An elastic component, a long transition (consequence of the beam plastic hinges being formed simultaneously) and a narrow yield plateau characterize the pushover curve.The plastic rotation capacity, qp, is defined as the total plastic rotation beyond which the connection strength starts to degrade below 80% [7]. This definition is different from that outlined in Section 9 (Appendix S) of the AISC Seismic Provisions [8, 10]. Using Eq. (2) derived by Uang and Fan [7], an estimate of the RBS plastic rotation capacity was found to be 0.037 rad:Fyf was substituted for Ry•Fy [8], where Ry is used to account for the differ- ence between the nominal and the expected yield strengths (Grade 50 steel, Fy=345 MPa and Ry =1.1 are used).3.Experimental programThe experimental set-up for studying the behavior of a connection was based on Fig. 6(a). Using the plastic displacement dp, plastic rotation gp, and plastic story drift angle qp shown in the figure, from geometry, it follows that:And: in which d and g include the elastic components. Approximations as above are used for large inelastic beam deformations. The diagram in Fig. 6(a) suggest that a sub assemblage with displacements controlled in the manner shown in Fig. 6(b) can represent the inelastic behavior of a typical beam in a CGMRF.The test set-up shown in Fig. 8 wasconstructed to develop the mechanism shown inFig. 6(a) and (b). The axial actuators wereattached to three 2438 mm × 1219 mm × 1219mm (8 ft × 4 ft × 4 ft) RC blocks. These blockswere tensioned to the laboratory floor by meansof twenty-four 32 mm diameter dywidag rods. This arrangement permitted replacement of the specimen after each test.Therefore, the force applied by the axial actuator, P, can be resolved into two or thogonal components, Paxial and Plateral. Since the inclination angle of the axial actuator does not exceed 3.0 , therefore Paxial is approximately equal to P [4]. However, the lateral component, Plateral, causes an additional moment at the beam-to column joint. If the axial actuators compress the specimen, then the lateral components will be adding to the lateral actuator forces, while if the axial actuators pull the specimen, the Plateral will be an opposing force to the lateral actuators. When the axial actuators undergoaxial actuators undergo a lateral displacement _, they cause an additional moment at the beam-to-column joint (P-△ effect). Therefore, the moment at the beam-to column joint is equal to:where H is the lateral forces, L is the arm, P is the axial force and _ is the lateral displacement.Four full-scale experiments of beam column connections were conducted.The member sizes and the results of tensile coupon tests are listed in Table 2All of the columns and beams were of A572 Grade 50 steel (Fy 344.5 MPa). The actual measured beam flange yield stress value was equal to 372 MPa (54 ksi), while the ultimate strength ranged from 502 MPa (72.8 ksi) to 543 MPa (78.7 ksi).Table 3 shows the values of the plastic moment for each specimen (based on measured tensile coupon data) at the full cross-section and at the reduced section at mid-length of the RBS cutout.The specimens were designated as specimen 1 through specimen 4. Test specimens details are shown in Fig. 9 through Fig. 12. The following features were utilized in the design of the beam–column connection:The use of RBS in beam flanges. A circular cutout was provided, as illustr- ated in Figs. 11 and 12. For all specimens, 30% of the beam flange width was removed. The cuts were made carefully, and then ground smooth in a direct- tion parallel to the beam flange to minimize notches.Use of a fully welded web connection. The connection between the beam web and the column flange was made with a complete joint penetration groove weld (CJP). AllCJP welds were performed according to AWS D1.1 Structural Welding CodeUse of two side plates welded with CJP to exterior sides of top and bottom beam flan- ges, from the face of the column flange to the beginning of the RBS, as shown in Figs. 11 and 12. The end of the side plate was smoothed to meet the beginning of the RBS. The side plates were welded with CJP with the column flanges. The side plate was added to increase the flexural capacity at the joint location, while the smooth transition was to reduce the stress raisers, which may initiate fracture.Two longitudinal stiffeners, 95 mm ×35 mm(3 3/4 in ×1 3/8 in), were welded with 12.7 mm(1/2 in) fillet weld at the middle height of the webas shown in Figs. 9 and 10. The stiffeners werewelded with CJP to column flanges.Removal of weld tabs at both the top and bottombeam flange groove welds. The weld tabs wereremoved to eliminate any potential notchesintroduced by the tabs or by weld discontinuitiesin the groove weld run out regions.Use of continuity plates with a thickness approximately equal to the beam flange thickness. One-inch thick continuity plates were used for all specimens.While the RBS is the most distinguishing feature of these test specimens, the longitudinal stiffener played an important role in delaying the formation of web local buckling and developing reliable connection performance4. Loading history Specimens were tested by applying cycles of alternated load with tip displacement increments of _y as shown in Table 4. The tip displacement of the beam was imposed by servo-controlled actuators 3 and 4. When the axial force was to be applied, actuators 1 and 2 were activated such that its force simulates the shear force in the link to be transferred to the beam. The variable axial force was increased up to 2800 kN (630 kip) at 0.5_y. After that, this lo- ad was maintained constant through the maximum lateral displacement.maximum lateral displacement. As the specimen was pushed back the axial force remained constant until 0.5 y and then started to decrease to zero as the specimen passed through the neutral position [4]. According to the upper bound for beam axial force as discussed in Section 2 of this paper, it was concluded that P=2800 kN (630 kip) is appropriate to investigate this case in RBS loading. The tests were continued until failure of the specimen, or until limitations of the test set-up were reached.5. ConclusionsBased on the observations made during the tests, and on the analysis of the instrumentation, the following conclusions were developed:1. The plastic rotation exceeded the 3% radians in all test specimens.2. Plastification of RBS developed in a stable manner.3. The overstrength ratios for the flexural strength of the test specimens were equal to 1.56 for specimen 1 and 1.51 for specimen4. The flexural strength capacity was based on the nominal yield strength and on the FEMA-273 beam–column equation. 4. Although flange local buckling did not cause an immediate degradation of strength, it did induce web local buckling5. The plastic local buckling of the bottom flange and the web was not accompanied by a significant deterioration in the load-carrying capacity.6. The longitudinal stiffener added in the middle of the beam web assisted in transferring the axial forces and in delaying the formation of web local buckling. How ever, this has caused a much higher overstrength ratio, which had a significant impact on the capacity design of the welded joints, panel zone and the column.7. A gradual strength reduction occurred after 0.015 to 0.02 rad of plastic rotation during negative cycles. No strength degradation was observed during positive cycles.8. Compression axial load under 0.0325Py does not affect substantially the connection deformation capacity.9. CGMRFS with properly designed and detailed RBS connections is a reliable system to resist earthquakes.弯钢框架结点在轴向变化荷载和侧向位移的作用下的周期性行为摘自：钢结构研究杂志。